1

ASSESSMENT OF SINGLE-STOREY PRECAST CONCRETE

INDUSTRIAL BUILDINGS WITH HINGED BEAM-COLUMN

CONNECTIONS WITH AND WITHOUT DOWELS

Manya DEYANOVA1, Stefano PAMPANIN2 and Roberto NASCIMBENE3

ABSTRACT

Recent seismic events in Europe and especially in Northern Italy have motivated a lot of analytical and

experimental research on the seismic behaviour of single-storey precast reinforced concrete (RC)

industrial buildings used mainly for manufacturing plants and storehouses. Seemingly with a regular

and straightforward load path, the seismic response of these buildings is characterised by many

uncertainties. One of the main source of uncertainty is the local behaviour of the semi-rigid beam-

column connections, also defined as pinned or hinged connections. Research by others has shown that

ignoring the rotational and ultimate capacity of these beam-column connections when analysing the

global response of the building can be misleading and not necessary on the conservative side (Bellotti

et al., 2009). Although various codes provide design guidelines on the maximum strength of the most

common beam-column connections, engineers face the difficulty to predict their performance and

expected deformation/rotation capacity when assessing existing precast RC structures. Another source

of ambiguity is the response of the slender columns, whose post-yielding behaviour and ultimate

capacity do not necessary follow well-known empirical models and procedures for columns with a much

lower shear aspect ratio (Fischinger et al., 2008). Moreover, the performance of such columns will be

significantly influenced by the variation of the axial load, P-Delta effects, insufficient transverse

reinforcement and, as recently proven by Boys et al., (2008), bidirectional loading effects.

The aim of the present study is to address the aforementioned issues through a detailed analysis

and application of existing theoretical methods. The case study is a single-storey industrial precast

building that consists of two prototype frames with the most common geometry and beam-column

connections found in this building typology in Italy. The selection was based on post-earthquake reports

after the Emilia-Romagna earthquake in Northern Italy, 2012 and technical documents issued by

RELUIS, 2008. The ultimate capacity, ductility and dissipative characteristics of the analysed frames

are discussed which is also useful for the development of adequate retrofit and design techniques.

INTRODUCTION

In the third quarter of the previous century, prefabricated RC structures started gaining popularity. Their

qualities were more often recognised and they became a common structural type where large open space

was required. Built 30 to 40 years ago, nowadays these structures, by definition, do not meet current

1 PhD student, ROSE Programme, UME School, IUSS Pavia, Italy, Institute for Advanced Study -

manya.deyanova@umeschool.it 2 Professor, Department of Civil and Natural Resources Engineering, University of Canterbury, Christchurch,

New Zealand - stefano.pampanin@cantebury.ac.nz 3 Researcher at European Centre for Training and Research in Earthquake Engineering (EUCENTRE), Pavia,

Italy - roberto.nascimbene@eucentre.it

2

seismic design requirements and yet, they are still operational even in regions with medium to high

seismicity. Interestingly enough, even precast RC buildings designed in the last decade, after the

introduction of capacity design and performance-based design concepts in the design code provisions,

seem to fall behind the constantly improving seismic design of conventional cast-in-place concrete

structures. Recent earthquakes in Italy (L’Aquila in 2009 and Emilia-Romagna in 2012) provided

unfortunate examples, which raised the urgent need for a comprehensive campaign of assessment and

retrofit of this type of structures.

Lately, extensive analytical and experimental investigations on connections, typical of the

European design practice of precast RC structures, has been carried out. The focus has been mainly on

the behaviour of beam-column connections with beams simply supported on the columns without

dowels (Magliulo et al., 2011) and the same with dowels (Capozzi et al., 2012; Aurelio et al., 2012;

Capozzi-Thesis, Zoubek et al., 2013; Fischinger et al., 2013; Kremmyda et al., 2013; Toniolo, 2013;

Belleri et al., 2012; Bournas et al., 2012; Psycharis et al., 2012). As a result, recommendations and

procedures for the calculation of the maximum shear capacity of a dowel beam-column connection are

available, with some of them already implemented in design guidelines and code provisions. Most of

these procedures, in terms of ultimate strength, show good agreement between each other and with

experimental data. As far as their monotonic and cyclic force-displacement response is considered,

however, the discrepancies are significant. One of the main reasons for this is the difficulty in the

prediction of the predominant failure mode of the connection.

This study focuses on the development of a methodology for the prediction of the response of two

typical beam-column connections, with and without dowels, for precast RC industrial buildings, by

analysing and comparing the variety of existing theoretical and experimental data on the topic. The

prototype connections were selected after the examination of numerous reports on the damages observed

after the Emilia-Romagna event in 2012 and of other published technical documents on precast industrial

buildings. The analytical non-linear response of the connections was then implemented in numerical

macro-models subjected to non-linear time-history analyses, so that issues like column slenderness and

geometrical non-linearity could also be considered. Knowing the distinctive features in the seismic

response of these types of precast structures is the key to adequate assessment, retrofit and design.

COMMON TYPES OF PRECAST RC STRUCTURES IN ITALY

Because Italy is one of the countries in Europe with the highest seismicity, information on the Italian

design and construction practice of precast RC structures is very helpful in pointing out their most

common performance deficiencies. For example, the post-earthquake reports (Belleri et al., 2014) after

the Emilia-Romagna events of May, 2012, showed the inadequacy of a number of existing beam-column

connections to withstand cyclic loading. The observed total collapse of the structures were mainly due

to the unseating of the beams in and out of the plane of the frames, coupled with minor cracking and

occasional spalling of the concrete at the base of the columns.

Considering only the single storey RC prefabricated industrial buildings, the data from these

reports was classified for this study and is presented in Fig. 1 to Fig. 4. Forty buildings were investigated

and their main properties in terms of span-length, column aspect ratio and beam-column connections

were summarised. Because the results do not meet the expectation for gradual improvement with time

of the seismic design concepts, the data collection was extended with a report by DPS/RELUIS

(Contegni et al., 2008) on different types of precast structures in Italy since the seventies. The report

was based on information about existing buildings provided by ASSOBETON – a consortium of several

construction companies in Italy.

Fig.1 to Fig. 4 show that there has not been any significant improvement in the seismic design of

the precast RC structures. More than 40% of the buildings were designed and constructed after 1996

when the code provisions in Italy started implementing capacity design and performance-based design

principles. Yet, the beam-column connections did not undergo conceptual changes. Despite the great

diversity of connections shown in Fig. 3 and Fig. 4, all of them are either with no lateral resistance or

with a very low ductility/deformation capacity and uncertain response under seismic excitation. For the

Emilia-Romagna region a reasonable explanation is that the area was not considered a seismic zone until

2003 (indicated in Fig. 1 with the vertical dashed line), however this does not apply to the buildings

M. Deyanova, S. Pampanin, R. Nascimbene 3

constructed after this period or to those from other seismic regions in Italy. Two types of connections

appear to be the most common: 1) a beam simply supported on a column through an elastomeric bearing

pad and 2) its improved version with one or two dowels protruding from the column and grouted in

ducts in the beam. These dowels may cross the beam height and be anchored at the top with nuts, or

cross only the bottom beam flange with no additional fastening but the grouting. The figures also show

that the most common beam span (Lb) is between 14m and 20m and the expected column aspect ratio

(height/width Lc/hc) is greater than 10.

Figure 1. Year of construction, beam span and column aspect ratio for 60 precast RC industrial buildings from

Italy: 40 from the Emilia-Romagna region and 20 from other regions

0

2

4

6

8

10

12

14

8<

L ≤

10

10<

L ≤

12

12<

L ≤

14

14<

L ≤

16

16<

L ≤

18

18<

L ≤

20

20<

L ≤

22

22<

L ≤

24

24<

L ≤

26

26<

L ≤

28

28<

L ≤

30

Nu

mb

er o

f b

uil

din

gs

Span Lb [m]

Most common span of main beams

Emilia-Romagna

RELUIS, 2008

0

2

4

6

8

10

12

14

5<

a ≤

7.5

7.5

< a

≤10

10<

a ≤

12.5

12.5

< a

≤15

15<

a ≤

17.5

17.5

< a

≤20

20<

a ≤

22.5

22.5

< a

≤25

Nu

mb

er o

f b

uil

din

gs

Aspect ratio Lc/hc

Most common column aspect ratio

Emilia-Romagna

RELUIS, 2008

18.2%

4.5%

18.2%

9.1%18.2%

22.7%

9.1%

Most common type of beams RELUIS, 2008

Tapered I Regural I ┴ - shaped

П - shaped H-Shaped Others64.4%

22.2%

13.3%

Most common type of beams Emilia-Romagna

Tapered I

Regular I

Others

Figure 2. Beam types for the same precast buildings as in Fig. 1

Rectangular

Others

Most common types of beams in Emilia-Romagna Most common types of beams after RELUIS, 2008

4

Figure 3. Most common beam-column connections from a database of 40 precast RC industrial buildings from

the Emilia-Romagna region in Italy

Figure 4. Most common beam-column connections from a database of 20 precast RC industrial buildings,

after RELUIS report (Contegni, Palermo, & Toniolo, 2008)

11.9%

45.2%

26.2%

9.5%

7.1%

Emilia-Romagna region, Italy

Centering pin

Simply seated on a fork-shaped column

Simply seated on a column

Seated on a column with dowels

Others

colu

mn

bea

m

elastomericpad 10-25mm

column

bea

m

colu

mn

bea

m

steel pin

elastomericpad 10-25mm

column

bea

m shallowgroove

150-2

00

beam

colu

mn

elastomericpad 10-25mm

1 (2) dowels

beam

colu

mn

elastomericpad 10-25mm

9.5%

38.1%

19.0%

19.0%

14.3%

RELUIS, 2008

Simply seated on a column

Seated on a column with dowels

П-beam on Џ-column

П-beam on ∩-column

Others

beam

colu

mn

elastomericpad 10-25mm

1 (2) dowels

beam

colu

mn

elastomericpad 10-25mm

grouteddowel

colu

mn

beam

neopren

steelplate

anchorscolumn

beam

M. Deyanova, S. Pampanin, R. Nascimbene 5

PROTOTYPE STRUCTURE

Based on the collected data, a case-study building designed according to the D.M. 1996 design code

(CS. LL. PP., 1996) and situated close to the epicentre of the Emilia-Romagna earthquake in 2012 was

selected. It combines the most typical structural features of precast RC single-storey warehouses in Italy

and is a good example of the damages that this building typology could experience under a real seismic

event. The building consists of cantilever columns forming frames in the transverse direction and no

additional lateral resisting system in the longitudinal direction except for the double-T roof panels,

which are expected to act as a diaphragm. The frames in the transverse direction (Frame Type A) consist

of I-shaped 18m long prestressed tapered beams, simply supported on the columns through an

elastomeric pad. In the perpendicular direction the last four grids are bridged by two main I-shaped

19.91m long prestressed beams connected to the columns with two dowels (Frame Type B). The frames

were investigated separately and their configuration is shown in Fig. 5.

Figure 5. Configuration of the analysed frames from the case-study building

RESPONSE OF THE SLENDER PRECAST COLUMNS

Typical for these type of structures, including the prototype one, is that the lateral resisting capacity

relies mainly on the cantilever columns, considered fixed into their socket foundations with plinths. As

mentioned already, these columns are expected to have different behaviour under horizontal loading,

compared to the most investigated and tested column types, to which the existing analysis methodologies

are calibrated. The main reason is the combination between slenderness and low transverse

reinforcement ratio. The aspect ratio, typically greater than 10, (in the investigated frames greater than

14) would suggest flexural failure as a predominant failure mechanism. At the same time, the low

amount of transverse reinforcement classifies these columns in the group of columns with expected

shear failure (Zhu et al., (2007)).

For better understanding of the subject different analysis methodologies were compared and are

presented in Fig. 6. The so called “DDBD (direct displacement-based design) procedure” is a bilinear

approximation and follows the recommendations for yielding curvature, ultimate curvature and plastic

hinge length by Priestley et al., (2007). “Cumbia” is a MatLab-based moment-curvature software

developed by the same authors, which provides information for different limit states. The limit states

are calculated based on material limits, as shown in Table. 1. The software also calculates the

displacement at expected buckling according to two methodologies: Moyer et al., (2003) (denoted as

M-K buckling), for which it assumes that the longitudinal reinforcement yields between stirrups; and

Berry et al., (2005) (denoted as B-E buckling). The “SeismoStruct” curves correspond to a fibre-element

model of the column with and without P-∆ effect, performed with the finite element software

SeismoStruct. The nominal strength of the concrete and the reinforcement are indicated in Table 2. The

axial load for each column is indicated with P in Fig. 6, together with the axial load ratio (ν=N/(Acf’cn))

and the ratio between the over turning moment due to P-Delta at yielding displacements and the yielding

moment capacity of the section (PΔy/My).

Evidently, the different methodologies agree well on the yielding displacement, which is also the

conclusion reached by Fischinger et al., (2008). As far as the ultimate displacement is considered,

however, they give wide range of results. It is worth noting that for the column with the highest axial

load the B-E buckling is very close to the point of abrupt drop in strength in the fibre-element model.

6

The ultimate drift of the columns was also evaluated following the approach by Haselton, (2006),

suggested by Fischinger et al., (2008) as the approach that yields better results among all the investigated

ones. The Haselton approach, however, yields unrealistic results for the prototype building, because of

the very low transverse reinforcement ratio for columns ρsh=0.04% compared to that of the structure

tested by Fischinger et al., (2008) - ρsh=0.86%.

Table 1. Material limit states considered by the MatLab algorithm "Cumbia"

after (Priestley, Calvi, & Kowalsky, 2007)

Limit State Concrete Strain Limits Steel Strain Limits

Serviceability 0.004 compression 0.015 tension

Damage control 2/3εcu 0.060 tension

In order to investigate the possibility of shear failure of the columns prior to flexural failure, the

drift at shear failure and the shear capacity were estimated according to Zhu et al., (2007), (Elwood et

al., (2008) and Priestley et al., (1994). All of these methodologies, however, consider columns with

aspect ratios less than 7, much less than the aspect ratio of the columns of the prototype building, thus

the results are either unreasonable, or should be critically treated. For instance, Table. 2 shows that the

method proposed by Zhu et al., (2007) leads to unrealistically large top displacements at shear failure,

due to the very high aspect ratio and at the same time the method by Elwood et al., (2008) yields much

lower displacements due to the very low transverse reinforcement ratio. The latter basically predicts that

these types of columns do not have any ductility, since they are expected to fail in shear almost straight

after yielding. This, however, has not been observed in any of the precast industrial structures

investigated after Emilia-Romagna events. An example is given in Fig. 7, which shows the damages in

one of the columns from the case-study building used for the prototype frames Type A and Type B. The

photographed crack pattern evidently corresponds to a flexure dominated response.

The last six columns in Table.2 calculate the shear capacity according to Priestley et al., (1994),

considering degradation with displacement ductility. The results show shear capacity much higher than

the expected shear demand in the columns, mainly due to the contribution of the high strength concrete.

Serv

iceabili

ty

Dam

age c

ontr

ol

Ultim

ateM

-K b

ucklin

g

B-E

bucklin

g

0

10

20

30

40

50

60

70

80

90

0 0.2 0.4 0.6 0.8 1 1.2

Forc

e [

kN

]

Displacement [m]

Frame Type A, external column

0.600

0.600

Ø24

Serv

iceabili

ty

Dam

age c

ontr

ol

Ultim

ate

M-K

bucklin

g

B-E

bucklin

g

0

10

20

30

40

50

60

70

80

90

0 0.2 0.4 0.6 0.8 1 1.2

Forc

e [

kN

]

Displacement [m]

Frame Type A, internal column

0.600

0.600

Ø20

Serv

iceabili

ty

Dam

age c

ontr

ol

Ultim

ate

M-K

bucklin

g

B-E

bucklin

g

0

10

20

30

40

50

60

70

80

90

0 0.2 0.4 0.6 0.8 1 1.2

Forc

e [

kN

]

Displacement [m]

Frame Type B, left column from Fig. 5

0.600

0.600

Ø20

Serv

iceabili

ty

Dam

age c

ontr

ol

Ultim

ate

M-K

bucklin

g

B-E

bucklin

g

0

10

20

30

40

50

60

70

80

90

0 0.2 0.4 0.6 0.8 1 1.2

Forc

e [

kN

]

Displacement [m]

Frame Type B, mid column from Fig. 5

0.600

0.600

Ø20

Displacement [m]

Cumbia SeismoStruct SeismoStruct with P-Delta DDBD procedure

Figure 6. Force-displacement curves for the top of the columns from the prototype building

P=370kN

ν=0.016

PΔy/My=12.5%

Stirrups ø8/0.2m Stirrups ø8/0.2m

P=665kN

ν=0.028

PΔy/My=26.3%

Stirrups ø8/0.2m

P=1070kN

ν=0.046

PΔy/My=37.1%

Stirrups ø8/0.2m

P=1480kN

ν=0.063

PΔy/My=46.8%

M. Deyanova, S. Pampanin, R. Nascimbene 7

Table 2. Comparison of the drift corresponding to shear failure and of the shear capacity of columns following

three methodologies

Figure 7. Crack pattern on one of the columns from the case-

study building, observed after Emilia-Romagna earthquake

The performance of the column with the highest axial load ratio was also numerically investigated

under the excitation of the second Emilia-Romagna event in May, 2012 from the station closest to the

case-study, direction North (Fig. 8). The non-linear time-history analyses were performed using the

same fibre-element model used for the push-over (Fig. 6 denoted as “SeismoStruct”), again with and

without P-Delta effects. Fig. 9 clearly shows that this type of columns is expected to behave elastically

for very high levels of drift - greater than 2%.

Figure 8. Elastic 5%-damped response spectra of the record from Emilia-Romagna event, May 29, 2012,

Mirandola (MRN) station, north

f'cn= 65 MPa

fyn= 473 MPa

ϕw

[mm]

s

[mm]n ρ'' θ

∆top

[m]θ

∆top

[m]

Vc

[kN]

Vn

[kN]

Vc

[kN]

Vn

[kN]

Frame Type A

external column10 0.36 0.6 370 67 8 200 2 0.0009 0.21 2.08 0.033 0.33 9 226 673 909 232 468

Frame Type A

internal column10 0.36 0.6 665 58 8 200 2 0.0009 0.21 2.06 0.032 0.32 16 226 673 916 232 475

Frame Type B

left column8.5 0.36 0.6 1070 68 8 200 2 0.0009 0.17 1.46 0.032 0.27 30 226 673 930 232 489

Frame Type B

mid column8.5 0.36 0.6 1480 80 8 200 2 0.0009 0.17 1.44 0.031 0.27 42 226 673 942 232 500

c - depth of compression zone Av - are of the transverse reinforcement

Vn - total shear capacity

μ∆ - displacement ductility

Zhu, et al., 2007

Elwood, et al., 2008

Priestley, et al., 1994

ρ'' - transverse reinforcement ratio

θ - drift at shear failure

∆top - displacement of the colunm top

Vp - contribution of the diagonal compression strut

Vs - contribution of the transverse reinforcement

Vc - contribution of the concrete

P - axial load

d - depth of the section

Ag - gross area of the section

L - column length

n - stirrup legs per section

s - spacing of stirrups

ϕw - diameter of stirrups

V - maximum shear demand

μ∆=3 k=0.1

Elwood, et

al., 2008

Zhu, et al.,

2007Vp

[kN]

Vs

[kN]

μ∆=1 k=0.29

Priestley, et al., 1994

ColumnV

[kN]

P

[kN]

d

[m]

Ag

[m2]

L

[m]

Ast

θ � 2.02ρ���0.025s

d� 0.013

L

d�

P

A�f��

θ �3

100� 4ρ�� �

1

40

ν

f ����

1

40

P

A�f′�

V� � k f′�0.8A�V� �d� c

2LPV �

A!f��d′

scot30°

0.00

2.00

4.00

6.00

8.00

10.00

0 1 2 3 4

a [

m/s

2]

Period [s]

0.00

0.20

0.40

0.60

0.80

1.00

1.20

1.40

0 1 2 3 4

v [

m/s

]

Period [s]

0.00

0.10

0.20

0.30

0.40

0 1 2 3 4

d [

m]

Period [s]

8

Figure 9. Non-linear time-history analysis of the column with the highest axial load ratio

Based on all the analyses on the response of slender precast RC columns the following

conclusions are reached: their ultimate displacements cannot be reliably predicted using theoretical

methods available in literature, developed for columns with much lower aspect ratio (typical for multi-

storey office and apartment buildings, or bridge piers); their response is expected to be elastic up to very

high drifts; their failure mechanism is likely to be flexural dominated even for columns with no capacity

design; shear failure might be expected after substantial damage in the plastic hinge region.

BEAM-COLUMN CONNECTIONS

The other main structural component of such systems is the beam-column connection. Two types of

connections were selected as the most common ones: 1) pinned connections with beams simply

supported on a column through an elastomeric bearing pad; 2) hinged connections with vertical dowels

protruding into the beam (see Fig. 3 and Fig. 4).

The first one relies only on the friction between the elastomeric pad and the concrete surface. As

the response is elastic-perfectly plastic, these types of connections theoretically respond with infinite

ductility. The beam seating length, however, is finite and the non-symmetrical accumulation of plastic

deformation in sliding under cyclic loading can lead to total collapse of the roof structure due to loss of

support. The shear resistance is formed by the elastic shear strength of the elastomeric pad until the level

of the friction force in sliding between the elastic material and concrete is reached and after that a

constant shear resistance in sliding. Through both stages, the shear capacity is proportional to the applied

gravity loads carried by the beam. For the cases of industrial building structures, with large spans, the

expected concentrated force at the beam support is significant and the contribution of the friction shear

resistance may not be negligible compared to the total lateral resistance of the connection. For the

prototype structure, the capacity of the beam-column connection in sliding was calculated following

several methodologies and a prediction for the force-displacement response is made based on a

combination of them. The comparison and the assumption (with a red line) for the Frame type A and B

are shown in Fig. 10. It is clear that the capacity of the beam-column connection in sliding is almost

50% of the shear capacity of a single column and, as will be seen later, more than 20% of the total

capacity of a dowel connection. This also implies that for the cases when the beam-column connection

relies only on friction, failure of the connection will occur before the plastic hinge at the base of the

column is activated.

-0.30

-0.20

-0.10

0.00

0.10

0.20

0.30

0.40

0 50 100

Top d

isp

lace

me

nt

[m]

Time [s]

without P-Delta

with P-Delta

-100

-50

0

50

100

-0.5 -0.4 -0.3 -0.2 -0.1 0.0 0.1 0.2 0.3 0.4 0.5

Ba

se s

hear

[kN

]

Top displacement [m]

without P-Delta

Bilinear approx.

with P-Delta

M. Deyanova, S. Pampanin, R. Nascimbene 9

Figure 10. Horizontal resistance of an elastomeric pad in contact with concrete.

Based on fib, (2008), Magliulo et al., (2011) and PCI design guidelines

The response of the second connection type combines the resistance in sliding with the resistance

of the dowels. A lot of analytical and experimental research has been conducted on the response of the

dowel connections. This includes the SAFECAST research project on mechanical connections in precast

structures under seismic conditions, conducted by a group of European associations of precast structures,

research centres and universities (Toniolo, 2013). As a result, there are many published methodologies

for the analysis of dowel connections. Fig. 11 compares some of them with the experimental results for

quasi-static cyclic tests on several configurations of dowel connections by Psycharis et al., (2012). The

calculations are performed with the geometry and materials corresponding to two of the tests, as shown

above each graph. It should be noted that the different theoretical procedures were calibrated to different

case-studies. The fib bulletine n.43, (2008) for example, considers one-sided dowels with no concrete

edge failure but only dowel failure due to deformations in the plastic hinge and crashing of the

surrounding concrete under to monotonic loading. The same refers to the research performed by

Capozzi, PhD thesis. On the contrary, Tanaka et al., (2011) and Capozzi et al., (2012) considered the

degrading dowel response after edge spalling, again under monotonic loading. Aurelio et al., (2012) and

Fischinger et al., (2013) tested double-sided dowels with no edge splitting. Some of the procedures were

calibrated to experimental set-ups with neoprene pad, whose contribution, however, has not been

explicitly differentiated.

The comparison in Fig. 11 shows that the maximum shear capacity predicted by the theoretical

methods agrees well between each other and with the experimental data. The discrepancies, however,

are significant for the force-displacement response up to failure. The main reason is that the failure of a

dowel connection is a complex combination of the dowel non-linear behaviour, its interaction with the

surrounding concrete and cracking and splitting of the concrete edges. It becomes even more

complicated to predict the predominant failure mode and the sequence of the non-linear stages, when

the dowel connects two concrete elements with different edge distances and/or different concrete

strength. An attempt was made to set up a methodology for the estimation of the dowel response based

on a combination of the compared procedures and two types of failure modes: failure mode dominated

by edge splitting (Fig. 11a), for which the response in “push” and “pull” directions is different after a

certain displacement; and failure dominated by large plastic deformations in the dowels (Fig. 11b), for

which the push-pull response is symmetric. The result of this methodology is presented with the black

lines, together with the exact values that build the curve.

0

10

20

30

40

50

60

70

80

90

0 10 20 30 40 50

Sh

ea

r re

sis

tan

ce [

kN

]

Lateral displacement [mm]

Frame Type A N=285kN

0

10

20

30

40

50

60

70

80

90

0 10 20 30 40 50

Sh

ea

r re

sista

nce

[k

N]

Lateral displacement [mm]

Frame Type B N=420kN

a = 0.24m

b = 0.42m

a = 0.15m

b = 0.42m

Elastic response Magliulo, et al., 2011 PCI fib 43, Fmax fib 43, Fmin Assumed

a

b

a

b

10

Figure 11. a)

Figure 11. b) Comparison between different analytical methodologies for the analysis of dowel connections with

experimental results.

Based on all the aforementioned observations, force-displacement and moment-rotation

relationships were developed for the modelling of the beam-column connection with a finite element

macro-model. The modelling schemes and the capacity curves are shown in Fig. 12 and Fig 13 for Frame

Type A and Frame Type B respectively.

fc'= 35 MPa

fy'= 580 MPa

0

50

100

150

200

250

300

350

0 5 10 15 20 25 30 35

Sh

ea

r re

sist

an

ce F

[k

N]

Lateral displacement ∆ [mm]

Comparison between theoretical and experimental response of dowel connection

dowel shear failure

front splitting

shear capacity

side splitting

∆=

0.1ϕ

F→

Ag

uia

r,2

01

1

∆=

0.2

6ϕ

F→

Ca

po

zzi,

20

12

fro

nt

split

∆≈

10m

m

∆≈

25

mm

pu

ll

F→

Cap

ozz

i,2

012

she

ar

cap

aci

ty

∆≈

30m

m p

ush

400

600

200

100

Ø25

push pull

fc'= 35 MPa

fy'= 580 MPa

156419.7287

0

20

40

60

80

100

120

140

0 5 10 15 20 25 30 35

Sh

ea

r re

sist

an

ce F

[k

N]

Lateral displacement ∆ [mm]

Comparison between theorwtical and experimental response of dowel connection

fib 43, 2008 Tanak, et al., 2011 Capozzi PhD Thesis

Fischinger, et al., 2013 Aguiar, et al., 2011 Capozzi, et al., 2012

Zoubek, et al., 2013 Toniolo, et al., 2013 Psycharis, et al., 2012 - Push

Psycharis, et al., 2013 - Pull Psycharis, et al., 2013 - theoretical Approximation

dowel shear failure

front splitting

shear capacity

side splitting

∆=

0.1ϕ

F→

Agu

iar,

201

1

∆=

0.1ϕ

F→

Ta

na

ka

,2

01

1

∆≈

10m

m

F→

Ca

po

zzi,

201

2

she

ar

cap

aci

ty

100

100

400

600

Ø16push pull

M. Deyanova, S. Pampanin, R. Nascimbene 11

Figure 12. Scheme of the finite-element model of the beam-column connection for Frame Type A and the

corresponding force-displacement relationship for the friction link element

NON-LINEAR TIME-HISTORY ANALYSIS

The non-linear response of the two types of beam-column connection is tested under the excitation of

the same record used for the test of the column dynamic response (see Fig. 8). A finite element model

of the connection is created, by modelling one full length column, half-span beams on both sides of the

connection and link elements as presented in Fig. 12 and Fig. 13. The column is modelled with a fibre

force-based frame element and the beams with elastic frame elements. The values of the masses and the

600

345

345

690

10

190 190

40

355

355

fibre force-based element

elastic frameelement

rigid frameelement

friction linkelement

gap linkelement

X

Z

140 160 140160

-60

-40

-20

0

20

40

60

-300 -200 -100 0 100 200 300

Fo

rce

[k

N]

Displacement [mm]

Z direction

X direction

unseating

-400

-300

-200

-100

0

100

200

300

400

-300 -200 -100 0 100 200 300

Fo

rce

[k

N]

Displacement [mm]

Frame Type B

Z direction

Y direction

unseating

dowel rupture

push pull

Z

X

150 150 150 150

600

750

750

1500

40

150 150

fibre force-based element

friction & rotationallink element

elastic frameelement

gap linkelement

rigid frameelement

-80

-60

-40

-20

0

20

40

60

80

-0.2 -0.15 -0.1 -0.05 0 0.05 0.1 0.15 0.2

Mo

me

nt

[kN

m]

Rotation [rad]

Around X

dowel

failure

beam-column

gap closing

Figure 13. Scheme of the finite-element model of the beam-column connection for Frame Type B and

the corresponding force-displacement and moment-rotation relationships for the link element

12

forces correspond to the tributary area of one beam span. An overview of the finite element model is

presented in Fig. 15.

Some results are presented in Fig. 14. The figure clearly shows that the global response and the

residual displacements are mainly influenced by the relative strength and capacity between the

connection and the adjacent elements. For example, for Frame Type A the base shear strength of the

column is ≈70kN without P-Delta effect and ≈40kN with P-Delta, and the friction capacity of the beam-

column connection is almost 30kN. This explains why when the geometrical non-linearity is not

considered, the relative displacement between the beam and the column top is significant, slightly over

0.12m. When P-Delta effects are considered, there is almost no relative displacement. The same refers

to Frame Type B, for which the strength of the connection is more than twice the shear of the column

that corresponds to the plastic hinge at the base. It should be kept in mind, however, that the numerical

tests performed consider a symmetrical set up, as far as geometry and loads are considered. In reality,

the different capacities of external and internal columns and pounding effects will introduce significant

irregularity and thus change the global response of the building.

Frame Type A Frame Type B

Figure 14. Results from non-linear time-history analyses of the beam-column connections

3.78E-07 0.000389 -2.12E-05 0.000389

4.97E-07 0.000509 -2.11E-05 0.000509

6.36E-07 0.000649 -2.09E-05 0.000649

7.95E-07 0.00081 -2.08E-05 0.00081

9.75E-07 0.000991 -2.06E-05 0.000991

1.18E-06 0.001192 -2.04E-05 0.001192

1.40E-06 0.001414 -2.01E-05 0.001414

1.64E-06 0.001656 -1.99E-05 0.001656

1.90E-06 0.001916 -1.96E-05 0.001916

2.18E-06 0.002194 -1.93E-05 0.002194

2.47E-06 0.002487 -1.90E-05 0.002487

2.78E-06 0.002795 -1.86E-05 0.002795

3.10E-06 0.003115 -1.83E-05 0.003115

3.44E-06 0.003444 -1.79E-05 0.003444

-0.16

-0.14

-0.12

-0.10

-0.08

-0.06

-0.04

-0.02

0.00

0.02

0 20 40 60 80 100

Dis

pla

cem

ent

[m]

Time [s]

Relative horizontal displacement beam-column top

Without P-Delta

With P-Delta

2.97E-07 0.000506 -7.74E-05 0.000506

4.06E-07 0.000685 -7.73E-05 0.000685

5.36E-07 0.000898 -7.72E-05 0.000898

6.89E-07 0.001149 -7.70E-05 0.001149

8.65E-07 0.001438 -7.69E-05 0.001438

1.07E-06 0.001767 -7.67E-05 0.001767

1.29E-06 0.002136 -7.64E-05 0.002136

1.54E-06 0.002545 -7.62E-05 0.002545

1.82E-06 0.002995 -7.59E-05 0.002995

2.12E-06 0.003485 -7.56E-05 0.003485

2.45E-06 0.004012 -7.53E-05 0.004012

2.79E-06 0.004577 -7.49E-05 0.004577

3.16E-06 0.005177 -7.45E-05 0.005177

3.55E-06 0.005809 -7.41E-05 0.005809

3.96E-06 0.00647 -7.37E-05 0.00647

-0.16

-0.14

-0.12

-0.10

-0.08

-0.06

-0.04

-0.02

0.00

0.02

0 20 40 60 80 100

Dis

pla

cem

ent

[m]

Time [s]

Relative horizontal displacement beam-column top

Without P-Delta

With P-Delta

0.25 2.16E-05

0.275 2.16E-05

0.3 2.16E-05

0.325 2.16E-05

0.35 2.15E-05

0.375 2.15E-05

0.4 2.15E-05

0.425 2.15E-05

0.45 2.15E-05

0.475 2.15E-05

0.5 2.14E-05

0.525 2.14E-05

0.55 2.14E-05

0.575 2.14E-05

-80

-60

-40

-20

0

20

40

60

80

-0.4 -0.2 0.0 0.2 0.4

Base

shear

[kN

]

Horizontal displacement [m]

Column top

Without P-Delta

With P-Delta

0.225 7.78E-05 3.00E-07 2.29E-05

0.25 7.78E-05 4.14E-07 -0.04515

0.275 7.78E-05 5.59E-07 0.005309

0.3 7.77E-05 7.08E-07 0.057604

0.325 7.77E-05 8.86E-07 -0.01977

0.35 7.77E-05 1.11E-06 0.012216

0.375 7.77E-05 1.35E-06 -0.00146

0.4 7.77E-05 1.60E-06 0.012865

0.425 7.77E-05 1.88E-06 -3.67E-03

0.45 7.77E-05 2.21E-06 -0.00024

0.475 7.77E-05 2.54E-06 0.001763

0.5 7.77E-05 2.90E-06 0.007163

0.525 7.77E-05 3.29E-06 0.004069

0.55 7.77E-05 3.69E-06 0.008807

0.575 7.77E-05 4.13E-06

-150

-100

-50

0

50

100

150

-0.4 -0.2 0.0 0.2 0.4

Base s

hear

[kN

]

Horizontal displacement [m]

Column top

Without P-Delta

With P-Delta

-2.13E-05 5.12E-05

-2.12E-05 0.001583

-2.11E-05 -0.00375

-2.09E-05 0.009592

-2.07E-05 -0.0005

-2.05E-05 0.00573

-2.03E-05 0.003241

-2.00E-05 0.001459

-1.98E-05 -7.69E-05

-1.95E-05 0.001241

-1.92E-05 0.002456

-1.88E-05 0.001649

-1.85E-05 0.002974

-1.81E-05 0.002087

-1.77E-05 0.002899

-80

-60

-40

-20

0

20

40

60

80

-0.4 -0.2 0.0 0.2 0.4 0.6

Base

shear

[kN

]

Horizontal displacement [m]

Beam at the connection

Without P-Delta

With P-Delta

-7.75E-05 2.29E-05

-7.73E-05 -0.04515

-7.72E-05 0.005309

-7.70E-05 0.057604

-7.68E-05 -0.01977

-7.66E-05 0.012216

-7.64E-05 -0.00146

-7.61E-05 0.012865

-7.58E-05 -3.67E-03

-7.55E-05 -0.00024

-7.52E-05 0.001763

-7.48E-05 0.007163

-7.44E-05 0.004069

-7.40E-05 0.008807

-7.36E-05 0.00287

-150

-100

-50

0

50

100

150

-0.4 -0.2 0.0 0.2 0.4

Ba

se s

hea

r [k

N]

Horizontal displacement [m]

Beam at the connection

Without P-Delta

With P-Delta

M. Deyanova, S. Pampanin, R. Nascimbene 13

Figure 15. Overview of the finite element model created in SeismoStruct for the analysis of the beam-column

connections

CONCLUSIONS AND FUTURE RESEARCH

This paper presents results of an investigation into the non-linear behaviour under seismic loading of

two types of pinned/hinged beam-column connections, typical of precast concrete industrial buildings

in Italy, together with the response of the slender columns. The following conclusions are reached:

o The beam-column connections of typical types of existing precast concrete structures do not

necessarily perform adequately under seismic loading. This is true even for those designed in

very recent times;

o The theoretical methods available in the literature and developed for columns with lower

aspect ratio cannot reliably predict the ultimate displacement of slender columns with aspect

ratio greater than 7;

o The response of slender columns like those investigated here can be expected to remain in

the elastic range for high levels of drift, up to 2%;

o The failure mechanism of the columns in this type of buildings is dominated by flexure even

for columns with no capacity design and low transverse reinforcement ratio;

o The existing theoretical approaches for the design of dowel connections agree well with each

other and with experimental results in terms of maximum design strength but not in terms of

force-displacement response up to failure. Based on them, in the present study a methodology

for the prediction of the cyclic response of dowel beam-column connections is proposed;

o The global response of the building is governed by the relative strength between the

connection and the column;

o If the beam-column connection is loaded symmetrically, it is expected to behave well under

seismic excitation. Different sources of irregularities, however, like different strength

between external and internal columns and pounding effects can significantly influence the

response of the connection. Further investigation is needed for the evaluation of these

phenomena;

o The influence of the rotation and uplift of the plinth foundations needs to be additionally

investigated;

o The conclusions presented in the present study can also be useful for the development of

adequate retrofit and design techniques.

14

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