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O C E A N I C S A U S T R A L I A

City of Port Adelaide Enfield

Volume 1 � Final Report

Port Adelaide Seawater Stormwater Flooding Study Principal Contacts Drew Jacobi (Tonkin Consulting) Bill Syme (WBM Oceanics Australia)

October 2005 Ref No 20020477RA3D

Table of Contents

City of Port Adelaide Enfield Port Adelaide Seawater Stormwater Flooding Study � Final Report 20020477RA3D.doc Revision: D Date: 10/10/05 Page: i


Table of Contents

City of Port Adelaide Enfield Port Adelaide Seawater Stormwater Flooding Study Final Report

VOLUME 1 � FINAL REPORT

1. Introduction 1

2. Land Subsidence 2 2.1 Background 2 2.2 Causes of Land Subsidence 3 2.3 Measurement of Land Subsidence 4 2.3.1 E&WS Survey Mark Re-levelling 4 2.3.2 Deep Bench Mark Survey 6 2.4 Future Changes to Land Subsidence Rates 6 2.4.1 Changes in Groundwater Use 6 2.4.2 Land Reclamation 7 2.5 Recommendations 7 2.5.1 Adoption of Land Subsidence Rates for this Study 7 2.5.2 Further Work 8

3. Tide Analysis 9 3.1 Storm Tide Analysis Summary 9 3.1.1 Overview 9 3.1.2 Tide Datums 9 3.1.3 Storm Height and Duration Analysis 10 3.1.4 Storm Tide Hydrograph Synthesis 11 3.1.5 Summary of Highest 25 Events 12 3.1.6 Basic Equations 14 3.2 Relationship between storm surges and rainfall events 14

4. Seawater Inundation and Protection 17 4.1 Introduction 17 4.2 Seawater Hydraulic Modelling 18 4.2.1 Data Collation and Review 18 4.2.2 Digital Elevation Model (DEM) Development 18 4.2.3 Hydraulic Model 19 4.3 Seawater Model Calibration 20 4.3.1 Selection of Seawater Calibration Event 20 4.3.2 Calibration to 1999 Event 23 4.4 Preliminary Future Conditions Modelling 26 4.5 Damages Assessment 27 4.6 Structural Condition Assessment 31 4.7 Seawater Flood Protection 31 4.8 Concept Sea Defence Upgrade 31 4.8.1 Background 31 4.8.2 Design Requirements 32 4.8.3 Concept Design 32 4.8.4 Outer Harbour 34 4.8.5 Estimated Costs 34

Table of Contents

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5. Local Stormwater Inundation and Protection 36 5.1 Introduction 36 5.2 Seawater � Stormwater Interaction Methodology 36 5.3 TUFLOW Model 37 5.4 ILSAX Model 38 5.5 Design �Average� Tide Cycle 39 5.6 100 year ARI Tide Cycle 39 5.7 Stormwater Drainage Performance Measurement 40 5.8 Stormwater Drainage Performance Requirements 40 5.9 Possible Future Increases to Rainfall Intensities and Impacts 41 5.9.1 Literature Review 41 5.9.2 Rainfall Intensity Changes Adopted for this Study 42 5.10 Anthony Street Catchment 42 5.10.1 Existing System 42 5.10.2 Proposed Upgrade Works 43 5.11 Centre Street and Jetty Road Catchments 43 5.11.1 Existing System 43 5.11.2 Proposed Upgrade Works 44 5.12 Hamilton Avenue Catchment 44 5.12.1 Existing System 44 5.12.2 Proposed Upgrade Works 44 5.13 Osborne Catchment 45 5.13.1 Existing System 45 5.13.2 Proposed Upgrade Works 45 5.14 Port Adelaide Centre Catchment 46 5.14.1 Existing System 46 5.14.2 Proposed Upgrade Works 46 5.15 Semaphore Road Catchment 47 5.15.1 Existing System 47 5.15.2 Proposed Upgrade Works 48 5.16 Future Sea Level / Land Subsidence Scenarios 48 5.17 Upgrade Costs 48 5.18 Recommended Future Scenario 49 5.19 Further Work 50

6. Wetlands and Ponding Basins Inundation and Protection 51 6.1 Methodology and Procedure for Wetlands and Ponding Basins Modelling 51 6.1.1 Background 51 6.1.2 Hydraulic Model Development 51 6.1.3 Digital Elevation Model Development 52 6.1.4 Boundary Conditions 52 6.2 Wetlands and Ponding Basins Existing Drainage Systems / Existing Development 52 6.2.1 Barker Inlet Wetlands Management Plan 52 6.2.2 Sea Level Rise / Land Subsidence Combinations 53 6.2.3 Design Wetland Standing Water Level 53 6.2.4 Concept Seawall Alignment 54 6.2.5 Storm Duration 54 6.2.6 Flooding Scenarios (Sensitivity Tests) 56 6.3 Peak Flood Levels 57 6.4 Further Work 59

7. Summary 60

8. References 62

Table of Contents

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Tables Table 2-1 Average Land Subsidence relative to Hope Valley Reservoir Bench Mark (Culver,

1970) 4 Table 2-2 Average Land Subsidence (1983-1994) relative to rock pin bench marks at the

Tollgate 6 Table 2-3 Discharge from Upper Tertiary Aquifer 6 Table 3-1 Tide Datums 9 Table 3-2 25 Highest Tidal Events 12 Table 3-3 Correlation Between Top 25 Storm Tide and Storm Surge Events 13 Table 3-4 Calculation Variables 14 Table 4-1 Dataset Acquisition and Sources 18 Table 4-2 Land Uses Types 19 Table 4-3 Meteorological Events with 25 Highest Peak Tide Heights 20 Table 4-4 Future Conditions Scenarios of Sea Level Rise and Land Subsidence 26 Table 4-5 Predicted Outer and Inner Harbour Levels 27 Table 4-6 RAM Damages for Large Non Residential Buildings 27 Table 4-7 Building Elevation Adjustments 28 Table 4-8 Building Size Classification 28 Table 4-9 100 year ARI Flood Damage Estimates 29 Table 4-10 Sea Defence System Required Threshold Levels 32 Table 4-11 Concept Sea Defence Upgrade Cost Estimates 34 Table 5-1 100 year ARI tide levels at Stormwater Outfalls 39 Table 5-2 Local Stormwater Drainage Upgrade Cost Estimates 49 Table 6-1 Design Standing Water Levels 53 Table 6-2 Wetland and Ponding Basins Peak Flood Levels 57 Table 6-3 100 Year Ponding Basin Peak Flood Levels Comparison 57 Table 6-4 Maximum Allowable Pond Flood Levels 58 Figures Figure 2-1 Bench Mark Level Changes relative to Hope Valley (Culver, 1970) 5 Figure 3-1 Rainfall versus Tidal Anomaly � Outer Harbour 16 Figure 3-2 Rainfall versus Tidal Anomaly � Outer Harbour 16 Figure 4-1 June 1999 Seawater Calibration Event (Levels to mOHD) 22 Figure 4-2 Preliminary Model Calibration Results 24 Figure 4-3 Revised Model Calibration Results 25 Figure 4-4 Flood Damage Costs (Lower Case) 30 Figure 4-5 Flood Damage Costs (Upper Case) 30 Figure 4-6 Demountable Flood Defence System (Demflood System) 33 Figure 6-1 Magazine Creek Wetland Outlet 100 Year ARI Water Level Comparison 54 Figure 6-2 Range Wetland Outlet 100 Year ARI Water Level Comparison 55 Figure 6-3 Ponding Basin Outlet 100 Year ARI Water Level Comparison 55 Figure 6-4 Barker Inlet Wetland Outlet 100 Year ARI Water Level Comparison 56 Appendices Appendix A Historical Upper Tertiary (T1) Aquifer Potentiometric Surface Levels (Gerges, 1996) Appendix B Storm Tide Analysis Additional Information Appendix C Historical Storm Tide Meteorological Profiles VOLUME 2 � DRAWINGS

Document History and Status

City of Port Adelaide Enfield Port Adelaide Seawater Stormwater Flooding Study � Final Report 20020477RA3D.doc Revision: D Date: 10/10/05 Page: iv


Document History and Status

Rev Description Author Rev�d App�d Date A Issued for review DJ/WJS KSS DJ 29/04/05

B Sec 3.2, 4.5 added DJ/WJS KSS DJ 18/05/05 C Final DJ/WJS KSS DJ 29/07/05 D Revised Ponding Basin Levels DJ/WJS KSS DJ 10/10/05

Introduction

City of Port Adelaide Enfield Port Adelaide Seawater Stormwater Flooding Study � Final Report 20020477RA3D.doc Revision: D Date: 10/10/05 Page: 1


1. Introduction

The City of Port Adelaide Enfield together with project partners including the Coast Protection Board, Flinders Ports, Torrens Catchment Water Management Board, Land Management Corporation and Transport SA have commissioned this Study to evaluate seawater and stormwater flood risks in Port Adelaide. Funding for the Study has also been obtained from the Commonwealth Government under the Natural Disasters Risk Management Studies Program. The aim of the Flood Risk Management Study is to identify the risks and develop and implement a strategy to protect the vulnerable areas of the City from the risk of seawater and stormwater flooding taking into account the possible sea level rise and land subsidence over the next hundred years. This Study is intended to be conducted in three phases as follows:

Phase 1: Risk Assessment/Preliminary Treatment � Analyse and evaluate risk of seawater and stormwater flooding and identify concept strategies.

Phase 2: Risk Treatment Study � Develop detailed strategies, including design and development of management measures and development controls.

Phase 3: Treatment Implementation � Implementation of control measures. This report examines Phase 1 of the Flood Risk Management Study.

Land Subsidence



2. Land Subsidence

2.1 Background

Historically land subsidence has been identified as a factor to be taken into account in the analysis of tidal records. It is of particular importance in the estimation of rates of sea level rise, and some studies have been undertaken to quantify land subsidence along the Adelaide coastline in order to allow proper interpretation of tidal records. Land subsidence was first identified as potentially occurring in Port Adelaide and the surrounding coastal regions as part of a Beach Erosion Assessment Study (Culver, 1970). This Study identified a trend in the change of Engineering and Water Supply Department bench mark levels across metropolitan Adelaide during the period of 1872 to 1969, relative to a reference bench mark associated with the Hope Valley reservoir that was assumed to be �stable�. A spatial variation across the 42 bench marks was observed, with calculated land subsidence rates increasing from nil in the CBD to 0.6 ft/century (1.8 mm/yr) in the Port Adelaide � Semaphore area. In order to provide greater certainty to the suggested land subsidence rates described above, particularly with respect to the stability of the reference bench mark, the then South Australian Coastal Management Branch performed a precise levelling survey. A network of deep benchmarks isolated from movement of the surface sediments were linked to �stable� rock pin benchmarks in the foothills by levelling to a high precision standard. Since the benchmarks were established in 1982, precise levelling surveys have been repeated in 1985, 1987 and 1994. The results from the relatively short time period over which measurements have been taken are in general agreement with the findings of the earlier Culver Study. Some additional levelling surveys have been undertaken using GPS however the vertical accuracy of this method is not considered to be adequate for this type of analysis, particularly over a short period of observation. More recently, land subsidence rate estimates were made from geological evidence including radiocarbon dated palaeosea level indicators (Belperio, 1993). Land subsidence rates ranging from 1.8 � 10 mm/yr were calculated for a number of areas in the region, with the major causes of the subsidence concluded to be surficial compaction associated with wetland reclamation and groundwater withdrawal from the upper Tertiary aquifer. Significantly, this work provides the only documented land subsidence rates for Gillman.

Land Subsidence



The discussion in this Section reviews the information provided by these investigations.

2.2 Causes of Land Subsidence

Land subsidence factors that have been identified as being relevant to the Port Adelaide region are discussed below. Groundwater withdrawal

Intensive groundwater pumping has occurred within the Port Adelaide region which has resulted in a significant modification to the potentiometric surface of the upper Tertiary (T1) aquifer. A major cone of depression has developed (refer Appendix A), attributable to the pumping regime at Penrice ICI and until recently, SAMCOR. This extraction began in 1957. The reduced water pressure within the aquifer allows for the slow drainage of clay and silt layers within or adjacent to the sand and gravel layers. Due to the compressibility of the clay and silt material compaction occurs which results in a lowering of the land surface. Land reclamation by draining of wetlands

The Gillman region has been subject to the artificial lowering (draining) of tidal wetland areas through the construction of levee banks and low-tide discharge sluice gates. This work commenced in 1894 and continued in stages to 1974. Levees and pond construction works drained the Gillman area in 1935. Lowering of the water table has caused consolidation, compaction and densification of exposed, previously saturated sediments, leading to land subsidence. In addition, portions of land reclaimed by the activities described above are underlain by Coastal Acid Sulphate Soils (CASS). The extent of soils with acid sulphate potential throughout South Australia was recently mapped (PIRSA, 2001). CASS occur in both the sulfide rich potential state and the oxidised activated actual state. Draining activities have resulted in the aeration of sulfide rich soils, allowing:

the previously saturated soil to dry out, consolidate and compact; microbial oxidation of organic matter resulting in mass loss (Stephens et al,

1984) oxidation of iron sulfide material to occur resulting in the production of

sulphuric acid. The acid has generated highly acid groundwaters in the Gillman area (pH 3.5-5.0) promoting extensive decalcification of sediments to depths greater than 2m within and beneath the mangrove peat layer

Land Subsidence



resulting in mass loss (Belperio, 1985; Belperio & Rice, 1989; Postma, 1983);

flushing of carbonate dissolution products as a result of the use of the reclaimed land as stormwater ponding basins (Belperio, 1993).

The land subsidence effects are most pronounced in areas where mangrove woodland has been cleared and the sediment subjected to the processes described above (Belperio, 1993). Land reclamation by filling

Extensive areas throughout the Study Area have been subjected to the placement of fill to achieve the current surface level. The weight of this overburden compresses underlying low density layers, in particular, the Holocene substrate which results in a lowering of the land surface. Current knowledge of the depth and extent of historical filling activities across the Study Area is fragmented and largely incomplete. Long term subsidence of St Vincent Basin

The Study Area is located within the St Vincent Basin. Long term subsidence rates within the basin are insignificant relative to the other factors described above.

2.3 Measurement of Land Subsidence

2.3.1 E&WS Survey Mark Re-levelling

The E&WS bench mark re-levelling data collected as part of a Beach Erosion Assessment Study in 1970, provides land subsidence statistics as summarised in Table 2-1 and shown spatially in Figure 2-1 below. Table 2-1 Average Land Subsidence relative to Hope Valley Reservoir Bench

Mark (Culver, 1970) Location Average Subsidence (mm/yr) Period Port Adelaide -1.55 1872 - 1969 Largs Bay -1.65 1890 - 1969 Largs Bay -1.83 1899 - 1969 Largs Bay -1.13 1899 - 1969 LeFevre Peninsula -5.67 1899 - 1969 Largs Bay -1.83 1899 - 1969 Largs Bay -1.71 1899 - 1969 Semaphore Park +0.21 1901 - 1969 Semaphore Park -2.19 1901 - 1969 Semaphore Park -0.18 1901 - 1969 Semaphore Park -1.80 1901 - 1969

Land Subsidence



Figure 2-1 Bench Mark Level Changes relative to Hope Valley (Culver, 1970) The following comments are made in relation to the land subsidence rates indicated by this Study:

The higher rates of land subsidence were found in the Port Adelaide / Semaphore region.

No land subsidence rates were estimated within the Gillman / Dry Creek region affected by the Port Adelaide wetland draining activities (completed in 1935).

Land Subsidence



The land subsidence rate at each location was calculated from two data points being the elevations of the PSM at the time of establishment and time of re-levelling. Hence the rate is an average which may not accurately describe the current rate of land subsidence. This is considered to be an important point as influencing factors such as groundwater pumping and draining of wetland areas commenced partway during the period over which the measurements were taken.

2.3.2 Deep Bench Mark Survey

The precise levelling data collected to date provides land subsidence statistics as summarised in Table 2-2 below. Table 2-2 Average Land Subsidence (1983-1994) relative to rock pin bench

marks at the Tollgate Location Average Subsidence (mm/yr) Outer Harbour 1.42 Port Adelaide 1.26 Fort Glanville 2.08 EWS Depot (Ottoway) 1.54

Another group of rock pin bench marks were established at Marino, however precise levelling surveys between the two locations indicate that Marino is rising at 1.0 mm/yr relative to the Tollgate. It has been suggested uplift at Marino may be due to tectonic forces including block tilting or fault movement (DEH, 2002). For this reason the subsidence rates relative to the Marino Bench Marks have been disregarded for the purposes of this Study. To date this work has not included levelling within the Gillman region. It is recommended that a station be established in this area to verify the increased local rates of land subsidence in this area reported to occur by the Belperio (1993) Study.

2.4 Future Changes to Land Subsidence Rates

2.4.1 Changes in Groundwater Use

Until recent times, the total extraction for the upper Tertiary (T1) aquifer within the Penrice � SAMCOR industrial zone (resulting in the cone of depression) was 2,688 ML/year (based on data collected from 1982-1984, Edwards et al, 1987) as summarised in Table 2-3 below. Table 2-3 Discharge from Upper Tertiary Aquifer

Location Approx Annual Discharge (ML) SAMCOR Gepps Cross 785

ICI Aust (Dry Creek) 145 Torrens Isle Quarantine 169

ICI Dry Creek 305

Land Subsidence



ICI Dry Creek 306 ICI Dry Creek 154 ICI Dry Creek 278 ICI Osbourne 546

TOTAL 2688 The closure of the South Australian Meat Corporation (SAMCOR) site in the late 1990s would have (based on 1982 data) reduced the total withdrawal to 1904 ML/year. There is insufficient data to indicate how this recent change in groundwater withdrawal regime will affect groundwater levels in the T1 aquifer and any consequential impact on land subsidence observed at the surface. However, it is expected that any changes (reductions) in observed rates would be slow and potentially insignificant.

2.4.2 Land Reclamation

Land reclamation activities (those associated with draining of the wetlands) were completed in 1970, however the duration over which land subsidence resulting from these activities will continue to occur is not known. It is expected that the rate of subsidence will reduce over time as the ongoing processes exhaust the potential for further subsidence, however monitoring of the levels in the Gillman region is required to confirm this.

2.5 Recommendations

2.5.1 Adoption of Land Subsidence Rates for this Study

Land subsidence rate(s) are required to be adopted for the assessment of seawater and stormwater flood risk. Assessments will be conducted for a number of future scenarios requiring the effects of land subsidence to be considered 50 and 100 years into the future. The Deep Bench Mark Study is considered to provide the most accurate and current representation of land subsidence rates within the Study Area. Over a 100 year period, they indicate that a total land subsidence in the range of 126 to 208 mm will occur over the Study Area. A single land subsidence rate of 2.1 mm/yr has been adopted over the Study Area for the purposes of this Study. While this over-estimates subsidence in some parts of the Study Area, it is considered appropriate to provide some conservatism (albeit small) particularly in risk assessment.

Land Subsidence



It is recognised that subsidence within the Gillman region is expected to be greater due to the factors described above in Section 2.2. However, uncertainty remains on the current rate (in the range between 2.1 � 10 mm/yr). In determining an appropriate rate to apply to this portion of the Study Area consideration needs to be given to the manner in which the rate would be applied. Where the subsidence rate is used to design permanent works that are incapable of being modified or altered to provide protection from flooding due to subsidence, a rate at the higher end of the range (10 mm/yr) should be adopted. Such works are likely to include land filling operations for the purposes of development. Further investigations are described in Section 2.5.2 below to provide a higher degree of certainty in relation to this rate. Where the rate is used in the design of works that are able to be altered, adoption of a lower rate, consistent with that over remainder of the Study Area is considered to be appropriate (2.1 mm/yr). Such works are likely to include seawater protection levees and structures. Implicit in this approach is a requirement for ongoing monitoring of the subsidence rates. In adopting a subsidence rate of 2.1 mm/yr over the Study Area, it is expected that the nature of works required within the Gillman area will be capable of later modification. The design of flood detention storages should be based on current surface levels (0.0 mm subsidence) as any subsidence will increase the volume available.

2.5.2 Further Work

It is recommended that further work be undertaken to improve the current level of understanding of land subsidence within Port Adelaide, as follows:

Undertake another precise levelling survey to complement the surveys conducted in 1985, 1987 and 1994;

Undertake re-levelling of Survey Marks (in co-ordination with DAIS Land Services Group) within the Gillman region such that movement of these marks since their establishment in the 1960�s can be assessed; and

Subject to the success of the above task, establish a deep bench mark within Gillman for future precise levelling surveys such that proper assessment can be made of high land subsidence rates within this region.

Tide Analysis



3. Tide Analysis

3.1 Storm Tide Analysis Summary

3.1.1 Overview

A statistical analysis was carried out to predict the 100 year storm tide level and duration off-shore of Barker Inlet. The findings from the analysis were used to synthesise a storm tide hydrograph (sea water levels versus time) for application to a hydrodynamic model that can predict over time seawater inundation of low lying areas around Port Adelaide. The analysis comprised the following tasks: 1. Collation and review of NTF Tidal Records; 2. The establishment of a relationship between the Inner and Outer harbours; 3. The replacement of gaps in the Outer Harbour record with Inner Harbour records; 4. An initial analysis of exceedence probabilities of Outer Harbour levels and

durations; 5. A review of the time-series of tidal anomalies (differences between recorded and

predicted, ie. an indicator of the occurrence of a storm surge) at the Outer Harbour;

6. Exceedence Probability Statistical Analysis of surge height (anomaly) and duration at the Outer Harbour;

7. Correlation of meteorological data with surge occurrences; 8. Completion of long-term Exceedence Probability Statistical Analysis of tide and

surge; 9. Correlation of long-term analysis with an initial analysis of tide levels; and 10. Finalisation of the method for deriving storm tide hydrographs. The above analysis was carried out through the processing of hourly records at both Inner and Outer Harbour along with predicted tide levels over approximately a 60 year period.

3.1.2 Tide Datums

Tide levels referred to in this document are to either Old Harbour Datum (OHD) or Australian Height Datum (AHD). The relationship between datums is presented in Table 3-1. Table 3-1 Tide Datums

Datum Conversion to AHD#

Description

OHD -1.723 Old Harbour Datum

Tide Analysis



LAT -1.452 Harbour or Chart Datum from 2001 onwards AHD n/a Australian Height Datum

# Add this number OHD or LAT to convert to AHD

3.1.3 Storm Height and Duration Analysis

Following the completion of Tasks 1, 2, 3, 4 and 5 the following features were identified:

The events with the highest tide heights all occurred between May and July. The peak of the meteorological events typically occurred in the morning at

8:00am. The pattern of the percentage of time in exceedence of tide height (PE) is

smooth and progressive essentially up to the highest recorded level of about 4.0mOHD (Refer to Drawing 3-1). This is true also for the residuals (i.e. recorded minus predicted tide) up to the peak value of about 2mOHD (Refer to Drawing 3-5).

There is an apparent anomaly in the mean duration data at about 3.5mOHD (Refer to Drawing 3-2). Review of the data indicates that this is associated with a reduction in the number of short duration events above that level combined with occurrence of some unusually long duration events (eg. 1981) above that level. Such an apparent anomalous pattern is not usually seen in nature and suggests that the data record is too short to contain sufficient events to directly represent the �average� condition.

The same pattern/anomaly is observed in the residuals (i.e. recorded minus predicted tide) at a value of about 1.5mOHD (Refer to Drawing 3-6).

Removal of some more extreme storm events from the dataset did not remove the anomaly of long duration events at elevated tide heights;

While some shorter period (i.e. ~ 1 hr) �wobbling� of the water levels is evident at the peak of some extreme events and is not well represented in the hourly digital data, this does not appear to affect the analysis results significantly.

From the above, the following conclusions and outcomes may be derived:

The relatively low number of events at elevated tide heights above about 3.5mOHD results in an apparent anomalous pattern of event durations and a lack of smooth progressive definition in the duration exceedence graphs at and above that level. Thus, the trend lines based on the data alone are not representative of the true behaviour of tide height and duration at elevated tide heights and require adoption of a form of extrapolation to the higher water levels as shown in Drawing 3-3.

Based on Drawing 3-1 and Drawing 3-3, ARI values for various levels and durations have been derived as depicted in Drawing 3-4.

Detailed duration data has also been analysed for the residuals, however, was not used further at this stage due to lack of a meaningful pattern for

Tide Analysis



extrapolation to the peak levels (Refer to Drawing 3-7). However, ARI values for the residuals in terms of level only (no duration dependence) have been derived from Drawing 3-5 and Drawing 3-6, as presented in Drawing 3-8.

At present, there is no evident explanation for this anomalous pattern other than the statistical occurrence of some unusually long duration events in the data record. However, this does not rule out the possibility of some real physical cause, as yet not identified, related to the way in which the storms propagate across from the west to the study area.

3.1.4 Storm Tide Hydrograph Synthesis

Drawing 3-4 has been used to define the peak of the 100 year ARI total storm tide hydrograph on the adopted basis of combining the 100 year corresponding water levels and durations. This yields both the peak value (approximately 4.05mOHD) and durations of levels down to about 3.0mOHD (~ 6 hrs). The lower parts of the storm tide hydrograph may be fitted to this via combining a typical tide and residual (storm surge) profile in such a way that it conforms with the defined peak. Total Storm Tide = Tide + Storm Surge In consideration that the 100 year ARI storm surge peak value is approximately 1.7m in depth (Refer to Drawing 3-8), it was found that the Mean High Water Spring Tide (MHWST) level of 2.4mOHD best fits the difference between the peak water level of the total storm tide and the storm surge component. The time-dependent shape of the 100 year ARI storm surge was formulated from consideration of a range of extreme historical events. Drawing 3-9 depicts the historical residual storm surges used, together with the adopted storm surge profile with a peak value of 1.65m in depth (close to the 100 year value of 1.7m) to be added to the MHWST to achieve the peak level of 4.05mOHD. Adopting a sinusoidal MHWST, the resulting storm tide hydrograph is as shown in Drawing 3-10. After discussions with Committee members and further analysis, a further 5 cm allowance was included to take into account long-term subsidence believed to have occurred at the gauge sites base on a rate of around 2 mm/year, giving an adopted 100 year storm tide peak of 4.10mOHD or 2.38mAHD. Appendix B includes additional information that contains an analysis by Flinders University on extreme seawater levels at Port Adelaide and a datum chart from Flinders Ports.

Tide Analysis



3.1.5 Summary of Highest 25 Events

Table 3-3 presents the above data in chronological order. Of interest, only three events occur within both lists indicating a low likelihood of peak tides and peak storm surges occurring at a similar time. Table 3-2 presents the 25 highest tidal events based on:

25 highest recorded peak tide levels (ie. tide plus surge); and 25 highest peak residual or anomaly heights (ie. surge component only

based on the recorded level minus predicted level) Table 3-2 25 Highest Tidal Events

Observed water levelsOrdered by date Ordered by peak water level reached

Peak Water Level Date/Time Peak Water Level Date/Time1 3.58 26/05/1933 10:00 1 3.57 12/07/1964 08:002 3.75 26/06/1945 08:00 2 3.58 26/05/1933 10:003 3.76 18/05/1953 8:00:00 AM (to 09:00) 3 3.59 11/06/1953 07:004 3.59 11/06/1953 07:00 4 3.59 23/06/1987 06:005 3.66 14/04/1956 8:00:00 AM (to 09:00) 5 3.6 05/08/1959 08:006 3.7 12/06/1956 08:00 6 3.6 28/05/1960 08:007 3.6 05/08/1959 08:00 7 3.61 27/07/1998 09:008 3.82 12/05/1960 08:00 8 3.601 30/05/1999 07:009 3.6 28/05/1960 08:00 9 3.62 29/09/1996 20:00

10 3.7 16/06/1961 08:00 10 3.66 14/04/1956 8:00:00 AM (to 09:00)11 3.57 12/07/1964 08:00 11 3.66 20/05/1977 08:0012 3.68 02/06/1965 08:00 12 3.68 02/06/1965 08:0013 3.75 28/06/1972 08:00 13 3.67 25/05/1994 07:0014 3.66 20/05/1977 08:00 14 3.68 23/06/1994 07:0015 3.85 01/06/1981 06:00 15 3.7 12/06/1956 08:0016 3.95 03/07/1981 08:00 16 3.7 16/06/1961 08:0017 3.59 23/06/1987 06:00 17 3.75 28/06/1972 08:0018 3.67 25/05/1994 07:00 18 3.75 26/06/1945 08:0019 3.68 23/06/1994 07:00 19 3.76 18/05/1953 8:00:00 AM (to 09:00)20 3.84 13/07/1995 07:00 20 3.819 13/06/1999 07:0021 3.62 29/09/1996 20:00 21 3.82 12/05/1960 08:0022 3.61 27/07/1998 09:00 22 3.84 13/07/1995 07:0023 3.601 30/05/1999 07:00 23 3.85 01/06/1981 06:0024 3.819 13/06/1999 07:00 24 3.872 21/06/2000 09:0025 3.872 21/06/2000 09:00 25 3.95 03/07/1981 08:00

Residual water levelsOrdered by date Ordered by peak water level reached

Residual Date/Time Residual Date/Time1 1.4 14/08/1934 16:00 1 1.26 9/06/1972 23:002 1.44 18/09/1937 3:00 2 1.26 10/06/1981 7:003 1.3 19/06/1943 14:00 3 1.26 30/08/1992 5:004 1.3 11/04/1948 3:00 4 1.28 30/08/1992 3:005 1.54 10/10/1948 18:00 5 1.3 19/06/1943 14:006 1.4 7/11/1948 7:00 6 1.3 11/04/1948 3:007 1.72 18/05/1953 6:00 7 1.3 5/06/1972 21:008 1.74 7/08/1955 18:00 8 1.3 25/03/1977 17:009 1.48 6/09/1957 3:00 9 1.3 1/06/1981 5:00

10 1.36 27/04/1960 3:00 10 1.3 5/08/1991 9:0011 1.34 31/07/1962 18:00 11 1.34 31/07/1962 18:0012 1.3 5/06/1972 21:00 12 1.34 10/06/1972 22:0013 1.26 9/06/1972 23:00 13 1.36 27/04/1960 3:0014 1.34 10/06/1972 22:00 14 1.4 14/08/1934 16:0015 1.52 2/12/1973 7:00 15 1.4 7/11/1948 7:0016 1.3 25/03/1977 17:00 16 1.44 18/09/1937 3:0017 1.3 1/06/1981 5:00 17 1.44 16/09/1999 6:0018 1.26 10/06/1981 7:00 18 1.46 23/06/1987 12:0019 1.54 26/03/1984 8:00 19 1.48 6/09/1957 3:0020 1.46 23/06/1987 12:00 20 1.48 12/09/1996 4:0021 1.3 5/08/1991 9:00 21 1.52 2/12/1973 7:0022 1.28 30/08/1992 3:00 22 1.54 10/10/1948 18:0023 1.26 30/08/1992 5:00 23 1.54 26/03/1984 8:0024 1.48 12/09/1996 4:00 24 1.72 18/05/1953 6:0025 1.44 16/09/1999 6:00 25 1.74 7/08/1955 18:00

Rosana Niven:N.B. Within 0.02mas 0.02m increments

L:\B14339.L.wjs\Tidal Analysis\Spreadsheets\Peak Events Summary - Observed and Residual 2003-03-08.xls

Tide Analysis



Table 3-3 Correlation Between Top 25 Storm Tide and Storm Surge Events Storm Events displayed in Chronological Order

Total Storm Tide Residual Storm SurgeDate/Time Date/Time

26/05/1933 10:0014/08/1934 16:0018/09/1937 03:0019/06/1943 14:00

26/06/1945 08:0011/04/1948 03:0010/10/1948 18:0007/11/1948 07:00

18/05/1953 8:00:00 AM (to 09:00) 18/05/1953 06:0011/06/1953 07:00

07/08/1955 18:0014/04/1956 8:00:00 AM (to 09:00)

12/06/1956 08:0006/09/1957 03:00

05/08/1959 08:0027/04/1960 03:00

12/05/1960 08:0028/05/1960 08:0016/06/1961 08:00

31/07/1962 18:0012/07/1964 08:0002/06/1965 08:00

05/06/1972 21:0009/06/1972 23:0010/06/1972 22:00

28/06/1972 08:0002/12/1973 07:0025/03/1977 17:00

20/05/1977 08:0001/06/1981 06:00 01/06/1981 05:00

10/06/1981 07:0003/07/1981 08:00

26/03/1984 08:0023/06/1987 06:00 23/06/1987 12:00

05/08/1991 09:0030/08/1992 03:0030/08/1992 05:00

25/05/1994 07:0023/06/1994 07:0013/07/1995 07:00

12/09/1996 04:0016/09/1999 06:00

29/09/1996 20:0027/07/1998 09:0030/05/1999 07:0013/06/1999 07:0021/06/2000 09:00

Tide Analysis



3.1.6 Basic Equations

The determination of the Annual Recurrence Interval (ARI) was based upon the general calculation method outlined below. The variables used within the calculation methodology are outlined in Table 3-4. Note that an �event� is defined as the exceedence of a particular water level from the time the water rises past that level to the time it falls below that level (event �duration�). Table 3-4 Calculation Variables

NH The hours per year in which the water level, h, is exceeded. NE The average number of events per year in which h is exceeded,

where the event may be comprised of several hourly exceedences. NN The number of events per year in which h is exceeded for at least

the specified duration. Dm The mean duration of events in exceedence of h. PE The percentage of the time in exceedence of h. PN The percentage of events in exceedence of h (NE) with a

corresponding duration of at least the specified duration. Where 1 year = 8760 hours

6.87H

E

NP

m

EE

D

PN

*6.87

Of the exceedence events a certain percentage will last for a specified duration.

m

NENEN

D

PPPNN

**876.0*

The Average Recurrence Interval (ARI) of the occurrence of an event in exceedence of a given water level for at least the given duration is the inverse of NN.

yearper Occurences ofNumber Average

1ARI

NE

m

PP

DARI

**876.0

3.2 Relationship between storm surges and rainfall events

Daily rainfall totals were analysed and compared with the tidal anomalies at both Outer Harbour and Inner Harbour gauges to ascertain whether there is a correlation between rainfall events and storm surge events. If a correlation can be reliably determined, this would be used to define the amount of catchment runoff that should

Tide Analysis



be modelled in conjunction with a storm tide event. Daily rainfall records from Rainfall Station 023036 (Pooraka) were obtained and compared with the long-term tidal anomalies as occurred at high tide. This translates to 37,680 comparisons (ie. the number of occurrences when a recorded high tide and daily rainfall record both exist) for the Outer Harbour gauge, and 28,139 for Inner Harbour. The tidal anomalies were analysed for different rainfall thresholds in increments of 1 mm. For each rainfall threshold, all the high tide anomalies that occurred on days when the rainfall threshold was equalled or exceeded were averaged and statistically analysed. Figure 3-1 and Figure 3-2 present the Rainfall Threshold plotted against the Tidal Anomaly for Outer and Inner Harbours. The blue diamonds represent individual observations, while the magenta line and squares is the mean high tide anomaly for each rainfall threshold. As can be seen, there is little correlation between rainfall and tidal anomaly indicating that there is no strong tendency, say, for rainfall to be greater when storm surges occur. On this basis, it was agreed that no reliable correlation between rainfall event probability and storm tide probability could be assumed.

Tide Analysis



Port Adelaide - Outer Harbour

-1500

-1000

-500

0

500

1000

1500

0 10 20 30 40 50 60 70 80 90 100

Rainfall Threshold (mm)

Tid

al A

no

mal

y (m

m)

Tidal Anomaly Mean Tidal Anomaly Figure 3-1 Rainfall versus Tidal Anomaly � Outer Harbour

Port Adelaide - Inner Harbour

-1500

-1000

-500

0

500

1000

1500

0 10 20 30 40 50 60 70 80 90 100

Rainfall Threshold (mm)

Tid

al A

no

mal

y (m

m)

Tidal Anomaly Mean Tidal Anomaly Figure 3-2 Rainfall versus Tidal Anomaly � Outer Harbour

Seawater Inundation and Protection



4. Seawater Inundation and Protection

4.1 Introduction

The Port Adelaide Enfield peninsular is located north-west of the city of Adelaide and lies adjacent to Gulf St Vincent and the Port River. The study area covers approximately 10,283 hectares with the majority of the area below the highest astronomical tide level. The mean spring tidal range of the area is between �1.15mAHD and 0.95mAHD. Drawing 4-1 shows the nominated study area for the seawater modelling. The banks of the Port River extend for approximately 50 kilometres and are protected in part by seawalls, embankments and floodgates. The land within the study area has experienced land subsidence as a result of numerous factors, which in combination with sea level rise, has resulted in a net land subsidence of 2.1mm per annum relative to current land levels. Under the Development Plan and State policy, all new development within the Port Adelaide Enfield region is currently required to be safe from 100 year storm floods and a rise in sea level of 0.3m. The policy also requires that new developments be adaptable to be safe to an additional sea level rise of 0.7m. Tonkin Consulting commissioned WBM to undertake seawater modelling of the Port Adelaide Enfield region to better understand the risks associated with seawater flood inundation within the Port Adelaide Enfield Council region. A two-dimensional (2D) TUFLOW hydraulic model of the region was developed to provide a base case against which management measures can be assessed.




4.2 Seawater Hydraulic Modelling

4.2.1 Data Collation and Review

As part of the assessment process WBM collected data required for the modelling component of this study. These datasets were predominantly acquired through the City of Port Adelaide Enfield, however, some data was acquired directly from other sources. Table 4-1 summarises the datasets collected as part of the assessment process. Table 4-1 Dataset Acquisition and Sources

Data Type Description Source

Drainage Port Adelaide Enfield Council pipes, pits and pumps network


Bathymetry Soundings of the Port River and

surrounding waterways Flinders Ports, MFP, City of Port Adelaide

Enfield

Tidal Records Outer and Inner Harbour observed tidal and residual historical records

National Tide Facility

Meteorological Data

Rain, wind and pressure gauge data Bureau of Meteorology

Significant Storm Records

Historic storm box records SA EPA

Topography Roads, spot heights, survey of seawalls and banks

Allsurv, Department of Land and Heritage SA

Aerial Photography

Outer Harbour region Aerometrex

Photogrammetry Topography taken from photogrammetry of the Port Adelaide Enfield council area


DTM Data City of Charles Sturt DTM data City of Charles Sturt

4.2.2 Digital Elevation Model (DEM) Development

A Digital Elevation Model (DEM) was developed for the purpose of providing necessary topographic and bathymetric data for the two-dimensional (2D) TUFLOW model. The development of the DEM was a staged process involving the incorporation of additional bathymetric and topographic data, in addition to the extension of the study area. The DEM was developed with a 5m grid cell resolution and was used as a basis for the development of a preliminary TUFLOW model of 50m cell size. Several areas of the preliminary DEM were identified that required further data (refer to Figure 2-1 in Discussion Paper 2, Port Adelaide Seawater and Stormwater Flooding Hydraulic Modelling). In particular, bathymetric data was required to further define the North Arm of the Port River channel.




A revised DEM was developed in order to incorporate the additional bathymetric and topographic datasets and the extension of the study area to include the City of Charles Sturt to the south. The revised DEM was developed with a 5m grid cell resolution and was used as a basis for the development of a higher resolution TUFLOW model of 30m cell size. Drawing 4-2 depicts the revised DEM developed for the study.

4.2.3 Hydraulic Model

A 2D/1D hydraulic model of the Port Adelaide Enfield region and Barker Inlet was developed using the flood and tide simulation software, TUFLOW (www.tuflow.com). The model includes a 1D network representing the major pipes and pumps. The 2D domain extends outwards from the Lefevre Peninsula to include Barker Inlet, Torrens Island and Gardens Island. Initially, the southern extents of the study area coincided with the southern boundary of the City of Port Adelaide Enfield Council. However, a revision of the study resulted in the extension of the study area to West Lakes in the City of Charles Sturt to the south. The boundary of the 2D domain is situated in deep water adjacent to Outer Harbour and represents the ocean behaviour in the inlet. The hydraulic model is based on a 30 m square grid. Each square grid cell contains information on ground topography sampled from the DEM at a 15m spacing, surface resistance to flow (Manning�s n value) and initial water level. Initially the hydraulic model was developed with a grid size of 50m, however, in order to achieve a greater resolution, the grid size was refined to 30m. Twelve areas of different land-use type were also established for the study area based on aerial photography. These land-use types were the basis for the assignment of Manning�s n values as shown in Table 4-2. Table 4-2 Land Uses Types Land Use Manning�s n Urban 0.2 Parks, Gardens and Vegetation 0.08 Commercial and Industrial 0.2 Sparse Terrestrial Vegetation and Grasses 0.06 Beach and Sand 0.04 Tidal Flats 0.05 Tidal Flats with some Vegetation 0.07 Mangroves and Intertidal Vegetation 0.09 Waterways 0.03 Main Channel (d/s) 0.025 Main Channel (u/s) 0.025 Barker Inlet 0.03




Drawing 4-3 shows the extent of the TUFLOW 2D/1D model developed for this assessment.

4.3 Seawater Model Calibration

4.3.1 Selection of Seawater Calibration Event

Sufficient historical data for use in model calibration was acquired in the form of sea level observations and residual water levels as far back as November 1940, and a list of significant meteorological events. A review of this historical data revealed that there was sea level information available for a range of meteorological events. Table 4-3 outlines the 25 meteorological events with the highest observed sea level.

Table 4-3 Meteorological Events with 25 Highest Peak Tide Heights

Number Date/Time (dd/MM/yyyy hh:mm)

Peak Tide Height Outer Harbour (mOHD)

1 03/07/1981 08:00 3.98 2 12/05/1960 08:00 3.90 3 01/06/1981 06:00 3.88 4 21/06/2000 09:00 3.86 5 26/06/1945 08:00 3.86 6 18/05/1953 08:00 3.85 7 13/07/1995 07:00 3.84 8 13/06/1999 07:00 3.82 9 28/06/1972 08:00 3.80 10 12/06/1956 08:00 3.78 11 16/06/1961 08:00 3.77 12 02/06/1965 08:00 3.74 13 14/04/1956 08:00 3.74 14 26/05/1933 10:00 3.72 15 20/05/1977 08:00 3.70 16 23/06/1994 07:00 3.69 17 25/05/1994 07:00 3.68 18 05/08/1959 08:00 3.68 19 11/06/1953 07:00 3.68 20 28/05/1960 08:00 3.67 21 10/05/1944 09:00 3.66 22 09/08/1955 00:00 3.65 23 12/07/1964 08:00 3.64 24 29/09/1996 20:00 3.62 25 03/08/1936 08:00 3.62

The full range of possible meteorological events were analysed in accordance with a selection criteria in order to select the best possible calibration event. The calibration event selection criteria is outlined as follows:

The quality of the historical water levels;




The height of the peak observed sea level; A single residual storm surge rather than several tidal anomalies in one

event; Coincident peaks of the residual storm surge and high tide; and Changes to the floodplain since the event.

Criteria assessment of the range of available calibration events resulted in a short-listing of 6 events as presented in Appendix C, from which the 13 June 1999 meteorological event was selected. The 1999 storm tide was created due to a low-pressure system off the coast of Port Adelaide. Sea levels peaked at approximately 3.82mOHD (2.10mAHD) at the Outer Harbour tide gauge. The peak of the residual storm surge aligns with the peak of the high tide and a single storm surge is observed over the length of the event. Analysis of the significant meteorological event rainfall record indicates that the 1999 meteorological event has the eighth highest recorded peak storm tide level. Figure 4-1 depicts the time-series data for the 1999 meteorological event. Of note is the tidal amplification that occurs at Inner Harbour.




Meteorological Event 13/06/1999

7:00am: 3.819m

0

0.5

1

1.5

2

2.5

3

3.5

4

4.5

12/06/1999 0:00 12/06/1999 12:00 13/06/1999 0:00 13/06/1999 12:00 14/06/1999 0:00 14/06/1999 12:00 15/06/1999 0:00

Date/Time

Wat

er L

evel

(m

Har

bo

ur

Dat

um

)

Outer Harbour Tide Gauge - Observed Inner Harbour Tide Gauge - Observed Predicted Tide

Figure 4-1 June 1999 Seawater Calibration Event (Levels to mOHD)




4.3.2 Calibration to 1999 Event

The calibration of the hydraulic models focussed on the recorded levels at the Inner Harbour tide gauge. Figure 4-2 and Figure 4-3 depict the calibration of the model to the Inner Harbour tide gauge for the peak of the 1999 meteorological event for the preliminary and revised models respectively. It can be seen from Figure 4-2 that good agreement is obtained between the recorded tidal record and the simulated seawater levels. The calibration is slightly less accurate for the revised model at the tide peak as indicated by Figure 4-3. A range of sensitivity tests were carried out using the preliminary and revised models. Conclusions from these tests were:

Different bathymetry, particularly in the mangrove areas and saltpans of Barker Inlet, was the main reason affecting the peak in the revised model. The preliminary model typically depicted these mangrove areas as being lower (0 to 0.9mAHD) than the revised model (0 to 2.0mAHD), and did not include many of the saltpans allowing greater tidal propagation and amplification into the Inlet. Until a complete bathymetric survey of Barker Inlet is available, this discrepancy will remain unresolved; and

Shifting the tidal boundary inwards has a slight affect in lowering the peak.




Port Adelaide Enfield Seawater Calibration - Preliminary Model

-1

-0.5

0

0.5

1

1.5

2

2.5

12/06/1999 0:00 12/06/1999 12:00 13/06/1999 0:00 13/06/1999 12:00 14/06/1999 0:00 14/06/1999 12:00 15/06/1999 0:00

Date/Time

Wat

er L

evel

(m

AH

D)

Outer Harbour Tide Gauge - Observed Inner Harbour Tide Gauge - Observed Inner Harbour Tide Gauge - Preliminary Model Calibration

Calibration_PA_1999-06-13_140_PO_2004-02-24.xls Figure 4-2 Preliminary Model Calibration Results




Port Adelaide Enfield Seawater Calibration - Revised Model

-1

-0.5

0

0.5

1

1.5

2

2.5

12/06/1999 0:00 12/06/1999 12:00 13/06/1999 0:00 13/06/1999 12:00 14/06/1999 0:00 14/06/1999 12:00 15/06/1999 0:00

Date/Time

Wat

er L

evel

(m

AH

D)

Outer Harbour Tide Gauge - Observed Inner Harbour Tide Gauge - Observed Inner Harbour Tide Gauge - Revised Model Calibration

Calibration_PA_1999-06-13_140_PO_2004-02-24.xls Figure 4-3 Revised Model Calibration Results




4.4 Preliminary Future Conditions Modelling

Preliminary modelling of future conditions was undertaken with the revised hydraulic model as per Section 4.2.1 (e) of the brief. The hydraulic model was also updated to include the Port River Expressway presently under construction. Table 4-4 outlines the future conditions presented in the study brief. Table 4-4 Future Conditions Scenarios of Sea Level Rise and Land

Subsidence Scenario Condition

Sea Level Rise (m)

Period of Land Subsidence

(years)

Description

S0 - - Existing case, no sea level rise or land subsidence

S1 0.30 50 Complies with current CPB requirements for infill development

S2 (not modelled)

0.10 100 Based on current IPCC projections for sea level rise over 100 years and using low end

value in the range plus 100 years of land subsidence

S3 0.50 100 As for Scenario 2 but using mid range value S4 0.88 100 As for Scenario 2 but using high end value

As per the study brief, Scenarios S1, S3 and S4 were simulated in the revised hydraulic model for the 100 year storm tide. Two cases, referred to as Lower Case and Upper Case, are presented for each Scenario. The Lower Case represents the situation where the non-tidal areas are assumed dry prior to the storm tide, while for Upper Case, low lying regions of non-tidal areas are assumed wet (due to higher mean sea levels, antecedent rainfall, etc) prior to the storm tide. The two cases were adopted because the extent of inundation is dependent on the available storage in the non-tidal areas, which, in turn, is dependent on the amount of ponded water in these areas (any ponded water reduces the available storage and causes greater inundation). The Lower Case (ie. non-tidal areas assumed dry) is considered to represent the greatest available storage, and therefore represents the case of less or lower inundation. For the Upper Case the starting water level was adopted as 0.68, 0.78, 1.18 and 1.56mAHD for Scenarios S0, S1, S3 and S4. The results of the modelling are presented in Drawing 4-4 (Lower Case) , Sheets 1 to 3, and Drawing 4-5 (Upper Case), Sheets 1 to 3. The drawings show the flood extent for Scenario S0 as flood depths (blue shades). They also show the incremental flood extent that is predicted to occur for Scenarios 1, 3 and 4 relative to the prior scenario (for example, the yellow shading shows the additional flood extent of S1 beyond that predicted for S0).




Table 4-5 presents the predicted levels at Inner Harbour for the modelled scenarios. Of interest is that the tidal amplification that occurs for Scenario S0 diminishes with increasing sea level rise. This is because increasing inundation of the floodplains causes increasing attenuation (dampening) of the storm tide peak. Should seawalls and/or landfill reduce the extent of the inundation, the tidal amplification is likely to correspondingly increase. Also of interest is there is little or no difference in the harbour due to the Lower and Upper Cases. Table 4-5 Predicted Outer and Inner Harbour Levels

Inner Harbour (mAHD) Scenario Outer Harbour (mAHD) Lower Case Upper Case

S0 2.39 2.50 2.50 S1 2.69 2.77 2.77 S3 2.89 2.90 2.90 S4 3.26 3.24 3.23

4.5 Damages Assessment

The assessment of damages was undertaken for the modelled future conditions scenarios listed in Table 4-4 for both the Lower and Upper Cases. Both RAM and ANUFLOOD methodologies were applied to calculate the equivalent flood damage estimations. RAM, the Rapid Appraisal Method for Floodplain Management (Rapid Appraisal Method (RAM) for Floodplain Management, Victorian Department of Natural Resources and Environment, Melbourne, 2000), recommends that the damage values include external and internal contents as well as structural damages. Thus, it recommends that the damage values should be applied to ALL inundated properties including those inundated above and below floor level. Damage calculations were completed for affected properties based on the following property classifications:

Residential property Commercial property Industrial property Garden equipment damages

Calculations based on survey data identified the actual damage to all buildings with an area less than 1000m2 to be equal to $16,400. For non-residential properties with an area greater than 1000m2, flood damage values were dependent on the size and condition of the property in question. Table 4-6 shows the calculated damages for large non-residential buildings. As no survey data for internal contents value was available, building condition was used to classify the level of internal contents. Table 4-6 RAM Damages for Large Non Residential Buildings Value of Contents Condition Damages per m2 Low Poor $36




Medium Average $64 High Good $160 (L:\B14339.L.wjs\Damages\Port_Ad_Building_Damages_s4.xls) ANUFLOOD is based on research conducted by the Natural Hazards Research Centre, and uses Stage-damage curves for flood damage assessment of residential and commercial areas. The methodology uses identical property classifications as for RAM, however, it also accounts for flood depth to calculate flood damage based on stage-damage curves, which relate damage to depth of inundation above the floor. Building locations are required for ANUFLOOD and RAM analysis. Using cadastral data the centroid for each property within the study area was located and assumed to be the building location (building footprints or centroids were not readily available). For the ANUFLOOD approach, which requires building floor levels for the stage-damage analysis, the elevation for each building was derived via the DEM (Digital Elevation Model) plus an adjustment between 0 and 1 metre above ground level. Table 4-7 identifies the building elevation adjustments for the suburbs within the area of study. Table 4-7 Building Elevation Adjustments Building Elevation Addition (m) Suburb 0.00 Largs Bay, North Haven, Peterhead 0.25 Birkenhead, Ethelton, Largs Bay, Largs North, North

Haven, Osborne, Port Adelaide, South Semaphore, Taperoo

0.50 Glanville, North Haven, Port Adelaide 1.00 Exeter, North Haven, Semaphore As survey data was unavailable for building size classification, property size was used to define the building size. Table 4-8 outlines the assumptions used for building size classification based on property size. Table 4-8 Building Size Classification Building Size Property Size (m2) Small >1000 Medium 1000< x >2500 Large <2500 Based on the building classification, size and floor level, both ANUFLOOD and RAM methodologies were used to calculate the estimated flood damage for the Port of Adelaide area. This was carried out for the 100yr events with a varying change in sea level, ranging between 0 and 0.88m based on IPCC projections for sea level rise over 100 years (ie. Scenarios S0, S1, S3 and S4).




The results of the flood damages assessment show that significant damages would occur, particularly for the longer-term future condition scenarios based on a rise in sea level. Figure 4-4 and Figure 4-5 present the damages estimates using the RAM and ANUFLOOD methods for the 100yr ARI events with varying sea levels and period of subsidence. Table 4-9 100 year ARI Flood Damage Estimates

ANUFLOOD RAM Scenario Condition Lower Case Upper Case Lower Case Upper Case S0 (Existing case, no sea level rise or land subsidence)

$7,900,000 $8,900,000 $26,000,000 $28,000,000

S1 (Complies with current CPB requirements for infill development, using a low end value sea level change and 50year period of land subsidence)

$54,000,000 $62,000,000 $65,000,000 $67,000,000

S3 (Based on current IPCC projections for sea level rise over 100 years using mid range value plus 100 years of land subsidence)

$112,000,000 $132,000,000 $90,000,000 $108,000,000

S4 (Based on current IPCC projections for sea level rise over 100 years using high range value plus 100 years of land subsidence)

$265,000,000 $310,000,000 $184,000,000 $200,000,000




$0

$50,000,000

$100,000,000

$150,000,000

$200,000,000

$250,000,000

$300,000,000

$350,000,000

S4 S3 S1 S0

100 Year ARI Flood Scenario

Co

st (

$)

ANUFLOOD RAM (L:\B14339.L.wjs\Damages\Port_Ad_Building_Sumaries.xls)

Figure 4-4 Flood Damage Costs (Lower Case) (L:\B14339.L.wjs\Damages\Port_Ad_Building_Sumaries.xls)

$0

$50,000,000

$100,000,000

$150,000,000

$200,000,000

$250,000,000

$300,000,000

$350,000,000

S4 S3 S1 S0

100 Year ARI Flood Scenario

Co

st (

$)

ANUFLOOD RAM (L:\B14339.L.wjs\Damages\Port_Ad_Building_Sumaries.xls)

Figure 4-5 Flood Damage Costs (Upper Case) (L:\B14339.L.wjs\Damages\Port_Ad_Building_Sumaries.xls)




4.6 Structural Condition Assessment

A visual inspection of the existing seawalls was undertaken in November 2003. Each unique reach of seawall was catalogued according to type and a qualitative condition rating and a photograph taken. The types of seawalls generally encountered including the following:

Sheet piling Concrete walls Riprap revetment Gabions Earth banks No man-made sea defence structure

The existing condition rating and type of sea defence structures are shown on Drawings 4-6 and 4-7.

4.7 Seawater Flood Protection

The level of seawater flood protection provided by the existing sea defence system was assessed by comparing the threshold overtopping level of the existing structures against peak high tide levels for a range of ARI. Outer Harbour peak tide levels were taken from the tide analysis (WBM, 2003a). A reach of seawall was assessed to provide a given level of protection if it has a threshold level that is above the corresponding tide level, with an additional:

100 mm to account for amplification of the tide from Outer Harbour to Inner Harbour, and

300 mm freeboard. This seawater flood protection standard provided by the existing sea defence system is shown on Drawing 4-8.

4.8 Concept Sea Defence Upgrade

4.8.1 Background

Strategies and concept designs have been developed to protect those areas shown to be subject to seawater inundation (refer to Section 4.4). Throughout the Study Area, a number of constraints exist that impact on the feasibility of each sea defence upgrade measure. These constraints include:

Wharf and other areas where direct access to the water is required Existing development close to the water Zones of existing development with historical significance




Areas that are environmentally sensitive, such as mangrove areas. Where possible, these areas have been delineated and are shown on Drawings 4-9, 4-10 and 4-11.

4.8.2 Design Requirements

The strategies have been developed to accommodate each of the four specified future sea level rise / land subsidence scenarios (refer Section 4.4). The required sea defence threshold level has therefore been set, for each scenario, to provide protection from the 100 year ARI tide event, with:

A base 2.50 mAHD threshold level An additional amount to account for sea level rise And additional amount to account for land subsidence (2.1 mm/yr) An additional 200 mm to account for amplification of the tide from Outer

Harbour to Inner Harbour, and 300 mm freeboard.

The threshold levels adopted for each scenario is summarised in Table 4-10 below. Table 4-10 Sea Defence System Required Threshold Levels

Scenario Condition

Required threshold level (mAHD)

1 3.405 2 3.310 3 3.710 4 4.090

4.8.3 Concept Design

The extent of works necessary in raising existing threshold levels to meet the design requirements described above in Section 4.8.2 was quantified by producing longitudinal profiles of the existing sea defence system, and superimposing the required threshold levels. Longitudinal profiles and location plans are presented in Drawing 4-12 Sheets 1-11 and Drawing 4-13 Sheets 1-4. The type of sea defence measure for individual reaches was selected based on the surrounding constraints. These measures include:

Earth Bank � Where sufficient space is available this is considered to be the cheapest option available.

Raised pavement levels at wharves � Activities in these areas rely on direct access the waterfront. The treatment proposed provides one way of




achieving a higher threshold level while still providing an adequate working area adjacent to the water.

Demountable Flood Defence at Road Crossings � In some locations, it is clear that vehicular access must be provided through the sea defence systems. Proprietary devices are available which provide one way of overcoming this problem. An example is shown in Figure 4-6 below.

Concrete wall � Limited space for bank batters is some locations will dictate that a vertical flood proof wall be adopted.

Riprap seawall � This applies particularly in locations where the existing riprap requires to be extended.

Raised bank with access track � This detail would apply to the bank separating the large ponding basins and North Arm.

Typical details for each of the measures specified are presented in Drawing 4-14 Sheets 1-3. Clearly these details would need to be customised to suit the constraints of each site.

Figure 4-6 Demountable Flood Defence System (Demflood System) At the Jervois Bridge, a barrier is proposed to be constructed across the Port River (a span of approximately 150 m). This is proposed due to the following:

Continuation of a raised seawall south of the Jervois Bridge would result in obstructing views from the recently developed Harbourside Quay residential area into the Port River, and

This alignment is shorter than the alternative alignment (continuing the seawall down to Bower Road) by approximately 1 km.




A range of flow regulation options could be incorporated to ensure that the Jervois Bridge barrier allows for adequate flushing of the southern water body and functionality of West Lakes. The form of the barrier requires further investigation to identify geotechnical conditions and operational requirements.

4.8.4 Outer Harbour

The sea defence concept design extends north along the eastern side of the LeFevre Peninsula to protect existing development. The storm tide inundation modelling has shown that for the S3 and S4 scenarios, portions of the Outer Harbor area east and north of Victoria Road are at risk of inundation. Further assessment is required to determine how future development in this region can manage risks associated with seawater and stormwater inundation. This assessment will need to determine the most effective solution from options including:

Site filling and conventional gravity stormwater drainage Seawalls and stormwater pump stations Stormwater detention basins Combinations of the above

4.8.5 Estimated Costs

Preliminary cost estimates have been prepared for the concept sea defence upgrade proposals. These estimates are prepared for general information only, and it is recommended that an appropriately qualified quantity surveyor be consulted to provide detailed advice regarding costs should more definitive estimates be required. It is expected that further development of the design concepts would be required to further refine the estimates. The estimates include a loading in addition to the construction cost to account for:

Survey and design (5%) Consultation and Negotiation (10%) Compensation (10%) Construction Administration (2.5%) Accommodation Works (30%) Contingencies (15%)

These estimates are presented in Table 4-11 below. The costs do not include GST. Table 4-11 Concept Sea Defence Upgrade Cost Estimates

Scenario 1 Scenario 2 Scenario 3 Scenario 4 Section 1 $13,000,000 $12,600,000 $14,400,000 $15,500,000 Section 2 $5,100,000 $4,900,000 $5,900,000 $6,700,000




Section 3 $2,800,000 $2,600,000 $3,500,000 $4,400,000 Section 4 $3,200,000 $3,200,000 $3,600,000 $3,900,000 Total $24,100,000 $23,300,000 $27,400,000 $30,500,000

Given the preliminary nature of the concepts and issues involved, we estimate the accuracy of the costs above to be +/- 30%. Special attention is drawn to the sensitivity of the cost estimates to the cost of supply of �environmentally clean� clay based fill for the creation of banks. Availability of this material is known to be variable over time due to significant demand for this type of material in the region. Sourcing clean fill is likely to require forward planning.

Local Stormwater Inundation and Protection



5. Local Stormwater Inundation and Protection

5.1 Introduction

A number of Council drainage systems have been identified for rigorous review and analysis to identify works required within these catchments to achieve an adequate level of protection against the combined seawater and stormwater flood risk. The catchments investigated by this Study include:

Anthony Street Osborne Road Hamilton Avenue Centre Street / Jetty Street Semaphore Road Port Adelaide Centre

These catchments are shown in Drawing 5-1.

5.2 Seawater � Stormwater Interaction Methodology

Within the Study Area, portions of land are low-lying to the extent that some areas are below recorded high tide levels. The performance of the stormwater drainage network within these catchments is affected by the prevailing downstream tide level in St Vincent Gulf and the Port River. The approach adopted for this Study has reflected the nature of each catchment investigated. An initial assessment found that for a number of the catchments under review, the performance of the stormwater drainage network was independent of tide level, due to:

A pump station draining a catchment to the Port River (Hamilton Avenue Catchment)

Catchment ground levels being sufficiently high to allow unimpeded gravity drainage (Anthony Street Catchment)

In other �small� catchments, the existing drainage standard of each drain reach was initially calculated, ignoring tidal influences on the operation of the gravity system. This found drainage standards to be generally low, ranging from less than 1 year ARI




to 5 year ARI. Consequently, there is very little latitude for tides to further reduce these standards. Hydrodynamic modelling of the stormwater networks, for a range of downstream tide cycle and rainfall event combinations, was found to be the most useful approach. This allowed for the existing network to be tested against the stated performance requirements (refer Section 5.8). This approach was successfully applied using the TUFLOW 1D / 2D model, which is described in more detail in Section 5.3. The significant advantage of this approach was that it allowed for assessment not only of the capacity of the underground drainage system, but also a rigorous assessment of the consequences of larger storm events. Many of the catchments in the Study Area are extremely flat (< 0.10% longitudinal road grade) and in many cases a single flood flow path was not clearly evident. In situations with numerous potential surface flood flow paths, a fully 2D approach is advantageous. However, the need to incorporate structures and the underground drainage network requires a link to a 1D system. TUFLOW allows a dynamic link (a water transfer every time-step and simultaneous simulation) between the 2D overland flow and the 1D underground drainage network. This 2D/1D approach provides a seamless link between the seawater and stormwater interface. A further advantage is that flood levels can be related directly to floor levels since the modelling package is GIS based, facilitating an efficient method for damage assessment. The link between the 1D pipe network and 2D surface model provided the advantage of allowing ponded flood flows to �fill� depressions in the surface automatically, without requiring individual height-storage relationships to be developed for each drainage node. As the surface model is seamless, these ponded volumes can spill and merge from one depression to another as dictated by the volume of floodwaters. Locations where non-performance occurred where readily identified from the mapping outputs.

5.3 TUFLOW Model

TUFLOW is a 2D/1D computer program for simulating depth-averaged, Two-dimensional Unsteady FLOW. The solution algorithm, based on Stelling (1984) and presented in Syme (1991), solves the full two-dimensional, depth averaged, momentum and continuity equations for free-surface flow. It was developed as part of a joint research and development project between WBM Oceanics Australia and The University of Queensland.




TUFLOW is specifically orientated towards establishing flow patterns in coastal waters, estuaries, rivers and floodplains where the flow patterns are essentially 2D in nature and cannot or would be awkward to represent using a 1D network model. Key steps to be undertaken in the development of a hydraulic model include:

development from the available bathymetrical data and topographic survey data of a detailed DTM over the domain of the fully 2D model. 2D grid sizes of 2.5m were created for these catchments;

digitising a MapInfo GIS layer from aerial photography that defines different vegetation / bed types (eg. mangroves, sand, marsh etc) for modelling the variation in flow resistance;

incorporate hydraulic controls such as levees (which are automatically modelled as broad-crested weirs) and any culverts, pump stations, floodgates which are modelled as 1D model inserts into the 2D model domain;

incorporation of a 1D model into the 2D domain that represents the piped drainage network of each of the small catchments;

establishment of boundary conditions including flow hydrographs to each drainage inlet (created by ILSAX) and a stage-time relationship of the tidal conditions at the outlets to the sea (from the seawater TUFLOW model).

5.4 ILSAX Model

Hydrological modelling using ILSAX was performed to determine peak flows occurring throughout each of the drainage systems for a range of average recurrence interval events. ILSAX is a rainfall-runoff routing program combining flows through a drainage network, and contains a procedure for calculating approximate pipe hydraulic capacities. Comparing peak flows against the estimated capacity of each drain reach assisted in the evaluation of existing standards. Subcatchments to each individual inlet were determined from field inspection and photogrammetry data. This field inspection process identified subcatchment modifications such as spoon drains and verified road gutter drainage patterns. The time of concentration for each individual subcatchment was calculated based on the gutter slope and maximum length of gutter flow to the inlet. This was in addition to the time of 5 minutes adopted for flow to travel from individual properties to the street water table. Data for each subcatchment was specified individually to represent the proportion of that area that is deemed to be impervious (e.g. rooves, paved areas). The remainder of the area was assumed to be pervious (e.g. grass, garden). The impervious area is further divided into impervious area with direct and indirect connection to the stormwater system.




Several sample blocks representative of residential areas within the different suburbs of the catchment were identified by inspection of aerial photography. The impervious area of each block was measured from aerial photography, and field inspection of the connection of downpipes to the street network enabled the directly and indirectly connected proportions to be calculated. Runoff coefficients of sample areas were checked for consistency against the range of values used for recent Urban Stormwater Master Plans and other hydrological studies undertaken in the region (Tonkin, 2003a, b, c, d, 2004) An initial loss of 45 mm and a continuing loss of 3 mm/hr from the rainfall hyetograph were used to determine the runoff from pervious areas (Kemp & Lipp, 2004).

5.5 Design �Average� Tide Cycle

The design �average� tide cycle adopted for this Study was the Mean High Water Springs (MHWS) Level. The level is the average high tide level produced during the full moon period (ie. this level is in the upper range of normally produced astronomical high tides). The Outer Harbour MHWS level of 0.95 mAHD was adopted throughout the Study. This level was adopted as a constant downstream level throughout the duration of the rain storm event. This was considered to be appropriate since these catchments had critical runoff response times of less than 1 hour, and the satisfactory performance of the minor drainage system during a simultaneous high tide level was of most interest.

5.6 100 year ARI Tide Cycle

The 100 year ARI tide cycle adopted for this Study utilised the calculated 100 year ARI peak tide levels (refer Section 3) as a peak tide cycle level. A sinusoidal function was used to define the tide cycle either side of the peak. 100 year ARI tide levels were extracted from the tide modelling results at each drainage outfall location as summarised in Table 5-1 below. Table 5-1 100 year ARI tide levels at Stormwater Outfalls

Catchment 100 year ARI peak tide level (mAHD) Anthony Street 2.38 Osbourne Road 2.38 Hamilton Avenue 2.48 Centre Street / Jetty Street 2.50 Semaphore Road 2.50 Port Adelaide Centre 2.50

These levels reflect the amplification of the tidal surge from Outer Harbor to Inner Harbor.




5.7 Stormwater Drainage Performance Measurement

A combination of the TUFLOW and ILSAX modelling results was used to assess the standard provided by the existing minor (underground) drainage system. This approach allowed a combination of indicators to be taken into account, including:

Surface ponding or surcharge from a drainage node (visible from TUFLOW output)

Capacity of a drain reach based on its size and grade The performance of the drain if the upstream system did not restrict the

peak flow rates passed downstream. Drawings have been prepared for each catchment displaying the assessed standard of each reach. Floodplain maps prepared from the TUFLOW modelling of each of the catchments provides comprehensive information with respect to the performance of the major (overland) drainage network for significant events.

5.8 Stormwater Drainage Performance Requirements

The minimum desirable stormwater drainage performance requirements, for a range of coincident rainfall and tide conditions, adopted for this Study are as follows:

Satisfactory performance of �Minor� System (no surcharging of underground drains, no exceedance of 5 year ARI design level of detention basins) for the:

Design �Average� Tide Cycle / 5 year ARI rainfall storm event 100 year ARI Tide Cycle / 1 year ARI rainfall storm event

Satisfactory performance of �Major� System (ponding within roadways

controlled to levels below adjacent floor levels, no exceedance of 100 year ARI design level of detention basins) for the:

Design �Average� Tide Cycle / 100 year ARI rainfall storm event Concept designs for proposed stormwater pump stations presented in the recommended upgrade works below have been sized and costed on the performance requirement of:

Pumps sized to convey the 5 year ARI flow (no storage necessary) Storage of some flows in excess of the 5 year ARI flow in a pipe storage

attached to the pump station Surface ponding of flows in excess of the capacity of the pump station and

the underground storage capacity, such that events up to the 100 year ARI event can be adequately managed.




Due to the preliminary nature of this investigation, further detailed analysis during the development of each pump station design would be required (as summarised in Section 5.19) to confirm an acceptable balance between the three inter-related variables described above.

5.9 Possible Future Increases to Rainfall Intensities and Impacts

5.9.1 Literature Review

Current knowledge of future rainfall characteristics is best described in the three reports described below. Third Assessment Report � Climate Change 2001 (IPCC, 2001) The Intergovernmental Panel on Climate Change (IPCC) has been established by the World Meteorological Organization (WMO) and the United Nations Environment Programme (UNEP) to �assess scientific, technical and socio- economic information relevant for the understanding of climate change, its potential impacts and options for adaptation and mitigation.� The report summarises the results of research undertaken by several hundred scientists worldwide over the preceding 5 years. The report presents future worldwide climatic predictions for a range of scenarios in relation to temperature, precipitation, extreme events and meteorological patterns, snow and ice coverage and sea level rise. These scenarios are linked to a range of greenhouse gas emission scenarios. Climate Change: An Australian Guide to the Science and Potential Impacts (Commonwealth of Australia, 2003) This report draws largely from the IPCC report but is set within the context of Australia. It also includes additional scientific developments and studies regarding climate change in Australia. Climate Change in South Australia (CSIRO, 2002) This report was prepared by the CSIRO Atmospheric Research Group (CAR) for the South Australian Government. Results from 11 climate models are presented, with predictions for changes to a number of key climatic variables given for eight regions within the State. The results of this report are considered to be of greatest relevance to this Study. Key outcomes for the Adelaide region are:

Average rainfall to decrease by 2-30% by 2070 Intensity of 10, 20 and 40 yr ARI events for the period 2011-2050 to

increase by 0-10% relative to the period 1961-2000.




5.9.2 Rainfall Intensity Changes Adopted for this Study

Rainfall patterns adopted for simulations of various rainfall events for the hydrological aspects of this Study are taken from Australian Rainfall and Runoff � Volume 2 (IE Aust, 1987). These rainfall patterns were developed from rainfall statistics collected prior to the time of publication and hence can be considered to be generally representative of the reference period 1961-2000 described above. Consideration of increased rainfall intensities is required as part of the planning and design of future stormwater flood protection works. This Study has taken into account the effects of increasing rainfall intensity runoff (10% increase) within one of three hydrological scenarios. These scenarios are:

Existing development, existing design rainfall Ultimate development , existing design rainfall Ultimate development, increased design rainfall

5.10 Anthony Street Catchment

5.10.1 Existing System

Currently, the Anthony Street Catchment collects stormwater from Military Road and discharges it to sea. This entire drainage system is well above the 100 year ARI tide level, and is therefore only affected by stormwater events. This drain system has less than 1 year ARI capacity, as presented on Drawing 5-2. Minor System

Design 5 Year ARI Storm The TUFLOW modelling results from the 5 year ARI storm (refer to Drawing 5-3) show that a portion of the flows in excess of the Anthony Street drain pond at the intersection of Anthony Street and Military Road. The majority of the flow passes down Windsor Street, Adelaide Street and Wills Street. Depths reach 0.3m to 0.5m in some locations. 1 Year ARI Storm Results from the 1 year ARI storm (refer to Drawings 5-4 Sheets 1-2 ) show that flows in excess of the Anthony Street drain travel to Wills Street. Ponding is generally restricted to the roadway, with a maximum depth of flooding between 0.1m and 0.2m. Major System

Design 100 Year ARI Storm TUFLOW results from the 100 year ARI storm (refer to Drawing 5-5) show a significant amount of ponding, both within the roadway and properties. Flood depths reach 0.5m to 1.0m in one residence on Wills Street.




5.10.2 Proposed Upgrade Works

Two branches are proposed to more effectively capture stormwater from Military Road. These branches extend from Anthony Street - south to Kalgoorlie Road and north to Musgrave Street. The main drain along Anthony Street also requires upgrade to dispose of the water captured from Military Road. Hydraulic grade line analyses of this system show that the proposed system is appropriate for all four sea level and land subsidence scenarios. Refer to Drawing 5-6 Sheets 1-3 for the proposed system, given the following three scenarios:

Existing Development, Existing Design Rainfall Existing Development, Increased Design Rainfall Ultimate Development, Increased Design Rainfall

5.11 Centre Street and Jetty Road Catchments


The Centre Street Catchment drain extends from Port River, back to Devon Street, while the drain for the Jetty Road Catchment is much smaller, extending to only to Victoria Road. Both the Centre Street and Jetty Road drains are gravity systems. Analysis of the existing stormwater systems within the Centre Street and Jetty Road Catchments (refer to Drawing 5-7) shows that they have less than a 1 year ARI capacity. Minor System

Design �Average� Tide Cycle with Design 5 Year ARI Storm Significant ponding occurs during the event of the 5 year ARI storm (refer to Drawing 5-8) in conjunction with the MHWS; with depths of 0.5 - 1.0m occurring in some locations. Design 100 Year ARI Tide Hydrograph with Design 1 Year ARI Storm Some ponding occurs in the event of a 1 year ARI storm in combination with the 100 year ARI tide hydrograph (refer to Drawings 5-9 Sheets 1-2). Modelling results show that most ponding is likely to be constrained to the roadway; however depths of 0.3 - 0.5m occur in some locations. Major System

MHWS with Design 100 Year ARI Storm The 100 year ARI storm in conjunction with the MHWS causes extensive ponding (refer to Drawing 5-10) throughout the Centre Street and Jetty Road Catchments. Ponding is not limited to the roadways, with many properties inundated; some with depths up to 0.5 - 1.0m.





A gravity system is not possible for the Centre Street and Jetty Road catchment as insufficient longitudinal grade can be achieved. It is proposed that the majority of the Centre Street system is diverted to connect to the Jetty Road system to reduce the extent of upgrade works within the Centre Street system (refer to Drawing 5-11 Sheets 1-3). Pumps are required for both proposed systems to discharge stormwater into the Port River.

5.12 Hamilton Avenue Catchment


The Hamilton Avenue system consists of a northern branch which extends back to Victoria Road and then to Brookman Street, and a southern branch along Bridges Avenue to Victoria Road. This stormwater system discharges to the Port River via a pump; consequently the tide does not influence its modelling. Currently, the Hamilton Avenue Catchment drainage system has less than 1 year ARI capacity, as shown on Drawing 5-12. Minor System

Design 5 Year ARI Storm The 5 year ARI storm (refer to Drawing 5-13) causes ponding in the roadways of Brookman Street, Bridges and Camilla Avenues, Mersey and Victoria Roads. Some ponding within properties is caused by this event. Design 1 Year ARI Storm Ponding caused by the 1 year ARI storm (refer to Drawing 5-14) is mainly constrained to the roadways of Brookman Street, Mersey Road and Bridges Avenue. Few residences are inundated by this event. Major System

Design 100 Year ARI Storm The 100 year ARI storm (refer to Drawing 5-15) causes extensive inundation within the Hamilton Avenue Catchment. The worst flooding is in residences on Bridges Avenue and Mersey Road. Depths reach 0.3 � 0.5m in some locations within the catchment. Ponding also occurs in locations between Victoria Road and Scott Street, between Hamilton Avenue and Marmora Terrace. Ponding also occurs north of Brookman Street.


The Hamilton Avenue pump system is at present drastically insufficient. The drains in this system also prove to be under capacity. It is proposed that the northern branch of the Hamilton Avenue drainage system is upgraded, as well as the pump system




(refer to Drawing 5-16 Sheets 1-3). It is proposed that the new pump station is located adjacent Bridges Avenue, rather than the existing pump station site adjacent Hamilton Avenue.

5.13 Osborne Catchment


The Osborne Catchment is bounded by Victoria Road, Lady Gowrie Drive and Marmora Terrace. Branches of the Osborne Catchment drainage system generally have 1, 5 or 10 year ARI capacity (refer to Drawing 5-17). Minor System

MHWS with Design 5 Year ARI Storm Virtually no ponding of stormwater occurs within the Osborne Catchment during the 5 year ARI event (refer Drawing 5-18) when it occurs in conjunction with the MHWS. Ponding is generally less than 0.1m; however, ponding does reach 0.3 � 0.5m in one location. Design 100 year ARI Tide Hydrograph with Design 1 Year ARI Storm Virtually no ponding occurs within the Osborne Catchment during the 1 year ARI storm event in combination with the 100 year tide hydrograph. Any ponding that does occur is limited to the roadways (refer to Drawings 5-19 Sheets 1-2). Major System

MHWS with Design 100 Year ARI Storm During the 100 year ARI storm event and the MHWS, some ponding occurs within the Osborne Catchment (refer to Drawing 5-20). The majority of ponding is constrained to the roadways however; some residences are inundated in this event. Depth of ponding reaches 0.5-1.0m in one location.


No modifications are proposed for the Osborne Catchment as ponding was found to remain within the roadway. Modelling of the above three seawater and stormwater event combinations has been repeated for �Scenario 3� (0.5m sea level rise and 100 years of land subsidence). Similar results were produced (refer to Drawings 5-21, 5-22 Sheets 1-2, 5-23) and hence no upgrade works are recommended in this catchment.




5.14 Port Adelaide Centre Catchment


The Port Adelaide Centre Catchment is serviced by several small stormwater systems. Currently, the stormwater systems within the Port Adelaide Centre Catchment generally have less than 1 year ARI capacity (refer to Drawing 5-24). Minor System

MHWS with Design 5 Year ARI Storm The combined 5 year ARI storm event and the MHWS cause a significant amount of ponding within the Port Adelaide Centre catchment (refer to Drawing 5-25). Ponding is not limited to the roadways, with several properties inundated. Flood depths reach 0.2 � 0.3m. Design 100 Year ARI Tide Hydrograph with Design 1 Year ARI Storm The 100 year tide hydrograph in conjunction with the 1 year ARI storm causes ponding through a significant number of properties with depths reaching 0.3 - 0.5m in some locations (refer to Drawings 5-26 Sheets 1-2). The majority of ponding however, is less than 0.1m deep. Major System

MHWS with Design 100 Year ARI Storm The 100 year ARI storm in combination with the MHWS causes significant flooding within the Port Adelaide Centre Catchment (refer to Drawing 5-27). Many properties are inundated due to this event. Ponding within roadways reaches depths of 0.3 - 0.5m.


One gravity system and two pumped systems are recommended for the Port Adelaide Centre Catchment to upgrade the existing situation (refer to Drawing 5-28 Sheets 1-2). Water from Santo Parade and St Vincent Street East is proposed to be re-directed to the Magazine Creek catchment by a gravity system. A pump on Timpson Street between Mempes Street and McLaren Road is proposed to discharge water collected from Divett, Todd, Timpson and St Vincent Streets to the Port River.

A pump on Mundy Street is proposed to discharge water collected from St Vincent and Nile Streets and McLaren Road to Port River.




The existing level of development within this catchment is considered to represent a state of �ultimate development�, with no significant increase in runoff expect to occur through redevelopment.

5.15 Semaphore Road Catchment


The drainage system for the Semaphore Road Catchment is composed of three individual systems which are located on Nelson Street, Fletcher Road and Semaphore Road. Currently, the drainage systems within the Semaphore Road Catchment have less than 1 year ARI capacity (refer to Drawing 5-29). Minor System

MHWS with Design 5 Year ARI Storm The MHWS in conjunction with the 5 year ARI storm event results in substantial flooding within the Semaphore Road Catchment (refer to Drawing 5-30). Flooding with depths up to 0.2 - 0.3m occurs on Hughes, Harris and Mead Streets. Flooding is not restricted to roadways, with several residences inundated in this event. A significant number of houses are inundated within the block bounded by Walker Street, Semaphore Road, Victoria Road and Heath Street. Another block that is affected by flooding is bounded by Hughes, Close and Moore Streets. Flooding also occurs within properties that flank the western side of railway and front Semaphore Road. Depths in these properties reach 0.1 - 0.2m. Design 100 Year ARI Tide Hydrograph with Design 1 Year ARI Storm The 1 year ARI storm in combination with the 100 year ARI tide hydrograph results in much ponding within roadways, as well as considerable ponding within properties between Close and May Streets (refer to Drawings 5-31 Sheets 1-2). Ponding within properties also occurs between Semaphore Road and Walker Street, and Semaphore Road between Mead and Teakle Streets. Ponding also occurs in the open spaces at the intersection of Fletcher Road and Heath Street and the intersection of Hughes and Close Streets. Major System

MHWS with Design 100 Year ARI Storm Severe flooding occurs in the Semaphore Road Catchment during the 100 year ARI storm in combination with the MHWS (refer to Drawing 5-32). The capacity of roadways is exceeded in most areas. Ponding reaches 0.3 � 0.5m in residential areas between Semaphore Road and Walker Street, Close and Roberts Streets, Close and May Streets, Roberts and Mead Streets as well as Semaphore Road between Mead and Wellington Streets. Significant ponding also occurs in the open spaces at the intersection of Hughes and Close Streets and Birkinhead Reserve.





Two pumped systems are proposed for the Semaphore Road Catchment to improve the present situation. One pump station is to be located on Victoria Road, the remaining pump system to be located on the reserve adjacent Le Fevre Peninsula Primary School. Proposed systems are presented in Drawing 5-33 Sheets 1-3, for the following three scenarios:

Existing Development, Existing Design Rainfall Existing Development, Increased Design Rainfall Ultimate Development, Increased Design Rainfall

5.16 Future Sea Level / Land Subsidence Scenarios

The effect of the four future scenarios (refer Section 4.4) has been considered in relation to the upgrade works within each catchment described above. Surprisingly, these scenarios were found to have no bearing on the options recommended for the catchments investigated as part of this Study, as discussed below. Anthony Street, Osborne Road Surface levels within the catchment are sufficiently high to allow for free gravity drainage under all 4 scenarios. Centre Street and Jetty Road, Port Adelaide Centre, Semaphore Road It was found that it is not feasible to provide gravity drainage with an acceptable level of performance within these catchments. Pump stations are recommended, which then allow the drainage of the catchment to perform independently of relative sea level. This assumes that an adequate sea defence system is in place. Hamilton Avenue This catchment is already served by a pump station and hence is similar to the catchments above. This result has significantly simplified the assessment of upgrade options required.

5.17 Upgrade Costs

Costings for the proposed drainage systems have been prepared. The estimated costs for the proposed drains include an allowance for construction costs, design, alteration of existing services, and contingencies of 15%. The costs do not include GST. These estimates are prepared for general information only, and it is recommended that an appropriately qualified quantity surveyor be consulted to provide detailed




advice regarding construction costs should more definitive estimates be required. It is expected that further development of the design concepts would be required to further refine the estimates. Table 5-2 Local Stormwater Drainage Upgrade Cost Estimates

Catchment Existing Development, Existing Design Rainfall

Ultimate Development, Existing Design Rainfall

Ultimate Development, Increased Design Rainfall

Anthony Street $720,000 $780,000 $1,120,000 Centre Street and Jetty Road - North

$880,000 $900,000 $1,370,000

Centre Street and Jetty Road - South

$2,890,000 $3,280,000 $5,170,000

Hamilton Avenue $1,390,000 $1,500,000 $2,260,000 Osborne $0 $0 $0 Port Adelaide Centre - East

$400,000 $400,000 $410,000

Port Adelaide Centre - Centre

$1,020,000 $1,020,000 $1,800,000

Port Adelaide Centre - West

$1,390,000 $1,390,000 $1,530,000

Semaphore - East $650,000 $850,000 $1,310,000 Semaphore - West $1,770,000 $1,910,000 $3,000,000 Total $11,100,000 $12,920,000 $17,950,000

The figures above demonstrate that adopting an �Ultimate Development� catchment status in the design of upgrade works results in a small cost increase (16%). Adopting an increased design rainfall scenario (associated with climate change) significantly increases (by 39%) the cost of upgrade works.

5.18 Recommended Future Scenario

Adoption of the �Ultimate Development, Existing Design Rainfall� scenario is recommended as a �middle of the road� approach. This recommended approach provides the flexibility to tolerate some increase in flows. Clearly some increase in runoff will be experienced, associated with new development. The �Increased Design Rainfall� scenario is a worst-case of current predictions of changes to rainfall patterns. Should an increasing trend in rainfall intensities be confirmed in the future, opportunities to manage this would include:

Changes to development requirements to more stringently control discharge and retention of stormwater




Increase the capacity of drainage infrastructure at the time of renewal or replacement

5.19 Further Work

It is recommended that detailed design development tasks be undertaken to progress the proposed upgrade works, as follows:

Confirm the suitability of the proposed drainage alignments, particularly with respect to services conflicts and negotiations / easement acquisition;

Undertake detailed modelling of the pump station upgrade proposals such that the pump and rising main design is sized sufficiently that an acceptable level of surface ponding occurs for the 100 year ARI event. It is anticipated that in some locations, there may be the need to consider a pump station with greater flow capacity;

Identify electricity supply and augmentation requirements and costs; Refine construction cost estimates based on the design development tasks

described above to support budgeting and funding application purposes.

Wetlands and Ponding Basins Inundation and Protection



6. Wetlands and Ponding Basins Inundation and Protection

6.1 Methodology and Procedure for Wetlands and Ponding Basins Modelling

6.1.1 Background

Stormwater runoff into the Barker Inlet, Range, and Magazine Creek wetlands, discharges either directly or via two large ponding basins into the Barker Inlet and Port River. The runoff process was analysed to determine the flood depths and heights for the 5 and 100 year ARI events. A multiple 2D domain hydrodynamic, TUFLOW model of the Port Adelaide Enfield wetlands and ponding basins was developed to analyse the seawater and stormwater interaction within the wetlands and ponding basins.

6.1.2 Hydraulic Model Development

The hydraulic model developed for the assessment of the wetland and ponding basin is an extension of the 30 m seawater flooding TUFLOW model described in Chapter 4. Four additional 2D domains of multiple cell size and orientation were linked to the seawater TUFLOW model to represent the complex conveyance and flood storage characteristics of the Port Adelaide Enfield wetlands and ponding basins. Finer cell sizes were used to more accurately analyse the flooding characteristics (eg. flood depths, direction and velocities of flow and possible break out locations) through the wetlands and basins as the 30 m resolution does not adequately depict the shape of key flowpaths in these areas. The Port Adelaide Enfield wetlands and retention basins have been sub-divided into the following 2D domains:

Magazine Creek Wetlands; Range Wetlands; Barker Inlet Wetlands; and Large ponding basins.

The grid cell size and orientation of each 2D domain has been defined based on the flowpath complexity and resolution required. The wetlands are based on 10m square cells at orientations derived from major flow paths and structures, while the large ponding basins are based on a 20m cell size. Each cell contains information on ground topography sampled from the DEM at a 5m spacing for the wetlands domains




and 10m for the basins domain, surface resistance to flow (Manning�s n value) and initial water level based on either mean seawater tide level or wetland design water level. Manning�s n values have been assigned based on the five areas of different land-use types identified in the seawater hydraulic modelling. The 2D domains are dynamically linked to the 1D network representing the major pipes/culvert network, as well as the larger 30m cell resolution seawater hydraulic TUFLOW model.

6.1.3 Digital Elevation Model Development

Ground topography is based on the revised DEM developed for the Seawater hydraulic model. Additional DEMs were developed to further define the Magazine Creek Wetland, Range Wetland, and the northern portion of the Barker Inlet Wetland using design drawings of the wetlands. The DEMs were developed with a 0.25m grid cell resolution.

6.1.4 Boundary Conditions

Boundary conditions for the model are sub-catchment inflows from the ILSAX modelling (carried out as separate studies), rainfall directly over the wetlands and ponding basins, and the tidal boundary as used for the seawater hydraulic modelling. Eight sub-catchment inflow hydrographs drain into the wetlands, with five directly into the Barker Inlet Wetlands, two into the Range Wetlands and a single inflow hydrograph into the Magazine Creek Wetlands. The timing of the ILSAX inflow hydrographs have been adjusted such that the peaks coincide with the peak of the tidal boundary. Drawing 6-1 presents the 2D/1D model developed for this assessment.

6.2 Wetlands and Ponding Basins Existing Drainage Systems / Existing Development

The hydraulic analysis has taken into account all relevant features in the three wetlands and the two large ponding basins. Gross Pollutant Traps have not been included in the model.

6.2.1 Barker Inlet Wetlands Management Plan

The relevant features and operation procedures from the Management Plan for the Barker Inlet Wetlands have been incorporated into the hydraulic analysis. The Management Plan states on page 27 and 65: �during storms, when storm tide surges or king tides may occur, all gates must be closed. ��




Storm events are often associated with storm and king tides, which may reach RL 3.00 AHD, which will prevent drainage of floodwaters through the sea wall. Under these conditions, it is imperative that the penstock gates on the tidal inlet culverts are closed while high tides persist. If they are not closed, tidal inflow combined with storm water inflow may raise water levels in the wetlands above the nominal maximum flood level, which may cause backwater flooding in the inlet drains and adjacent properties. High tides associated with storm events may occur at anytime throughout the year but especially during winter and spring (July � November). Generally, during these periods and for tides forecast above 2.5m, the penstock gates on the tidal inlet culverts should be set as follows, two closed and two half open.� The Barker Inlet Wetlands Tidal Culverts have been modelled as uni-directional channels in the 1D domain of the hydraulic model. Seawater is unable to flow from the sea into the wetlands at high tides, simulating closed penstocks. When the tide subsides, stormwater is able to drain from the wetlands out to sea, simulating the re-opening of the penstocks.

6.2.2 Sea Level Rise / Land Subsidence Combinations

The tidal boundary used for the wetlands and ponding basins modelling is based on future Scenarios S3 (MHWS Tide Event) and S4 (100 year ARI tide event), however no land subsidence has been modelled, in accordance with the approach described in Section 2.5.1. The wetland and ponding basin results produced for the S3 MHWS tide event are considered to be the appropriate results to adopt for further use. The S4 100 year ARI tide results are also presented to demonstrate the sensitivity of the adopted tide assumption on wetland and ponding basin peak flood levels.

6.2.3 Design Wetland Standing Water Level

Initial water levels modelled for the three wetlands are based on design standing water levels, as presented in Table 6-1. Table 6-1 Design Standing Water Levels Wetland Design Standing Water Level Magazine Creek Wetland -0.6 m AHD Range Wetland 0.3 m AHD Barker Inlet Wetland � Freshwater � Saltwater Intrusion Area

0.2 m AHD 0.0 m AHD (mean tide level)




The large ponding basins have been modelled initially as �empty�.

6.2.4 Concept Seawall Alignment

For the purpose of the wetlands and ponding basins inundation and protection modelling, the seawalls have been upgraded based on the proposed seawall alignment, thus preventing seawater overtopping into the basins.

6.2.5 Storm Duration

Sub-catchment inflows from the ILSAX modelling were provided for the 100 year and 5 year ARI flood events, for the 18, 24, 30, 36, 48 and 72 hours, storm duration, which were simulated through the TUFLOW model. Figure 6-1 to Figure 6-4 show water level time series at various locations within the wetlands and ponding basins for the different durations (against the S4 100 year ARI tide condition).

Magazine Creek Wetland Outlet

-1.5

-1

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Figure 6-1 Magazine Creek Wetland Outlet 100 Year ARI Water Level

Comparison




Range Wetland Outlet

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Figure 6-2 Range Wetland Outlet 100 Year ARI Water Level Comparison

Ponding Basin Outlet

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Figure 6-3 Ponding Basin Outlet 100 Year ARI Water Level Comparison




Barker Inlet Wetland Outlet

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Figure 6-4 Barker Inlet Wetland Outlet 100 Year ARI Water Level Comparison Peak flood levels are dominated by the 30 hour storm duration at all locations, as illustrated in the previous water level hydrographs. Drawing 6-2 presents the peak flood depths for the 100 year, 30 hour duration event, and Drawing 6-3 shows the 5 year, 30 hour duration event. The floodwaters remain contained within the wetlands and basins, even with the high ocean conditions (Scenario S4). Essentially, the wetlands and basins are able to absorb all of the floodwaters due to their large storage volumes. Clearly, this is dependent on the wetlands and basins being at their operating levels or empty prior to the flood event. A second round of TUFLOW modelling was undertaken, adopting the S3 MHWS downstream tide condition. Drawing 6-4 presents the peak flood depths for the 100 year, 30 hour duration event, and Drawing 6-5 shows the 5 year, 30 hour duration event.

6.2.6 Flooding Scenarios (Sensitivity Tests)

The 5 year and 100 year ARI flood events were also simulated for the following sensitivity tests. The hydrographs for these tests were generated from the ILSAX modelling carried out as a separate study.

Sensitivity 1 - Ultimate Catchment Development, Existing Rainfall Intensity




Sensitivity 2 - Ultimate Catchment Development, Increased Rainfall Intensity

Drawings 6-2 to 6-5 also show the peak flood levels at various locations within the wetlands and ponding basins. The existing case is shown as a red number, the ultimate catchment / existing rainfall as a blue number and ultimate catchment / increased rainfall as a green number. As can be seen, the sensitivity tests show an increase in flood levels (by up to 0.25m), although the floodwaters remain well contained within the wetlands and basins.

6.3 Peak Flood Levels

5 year and 100 year ARI peak flood levels recorded within each of the wetland and ponding basin regions for all sensitivity scenarios are summarised below in Table 6-2. Table 6-2 Wetland and Ponding Basins Peak Flood Levels

Location Existing Sensitivity 1 Sensitivity 2 5 yr 100 yr 5 yr 100 yr 5 yr 100 yr Barker Inlet Wetland - Inlet 1.49 2.12 1.52 2.30 1.53 2.34 Barker Inlet Wetland - Salisbury Hwy

1.10 1.83 1.13 2.02 1.16 2.06

Barker Inlet Wetland - Outlet 1.07 1.82 1.11 2.01 1.14 2.06 The Range Wetland - North Arm Drain Inlet

0.76 1.15 0.92 1.19 0.94 1.19

The Range Wetland - Hanson Road Drain Inlet

0.74 1.12 0.80 1.15 0.82 1.16

The Range Wetland - Outlet 0.74 1.11 0.80 1.14 0.81 1.14 Magazine Creek Wetland - Inlet

0.05 0.60 0.11 0.66 0.13 0.68

Magazine Creek Wetland - Outlet

0.05 0.60 0.11 0.66 0.13 0.68

Southern Ponding Basin - Culvert to Northern Basin

0.61 0.89 0.64 0.91 0.64 0.92

Northern Ponding Basin - Tidal Outlet Structure

0.05 0.60 0.11 0.66 0.13 0.67

The 100 year ARI peak pond levels in the three wetland / ponding basin areas compare to previous assessments as summarised in Table 6-3 below. Table 6-3 100 Year Ponding Basin Peak Flood Levels Comparison

Source Barker Inlet The Range Magazine Creek Tonkin & WBM (2005) 1.82 - 2.12 1.12 0.51-0.60 CMPS&F (1996) - 1.54 0.68 Woodward Clyde, AGC (1994) 1.70 - - BC Tonkin & Associates (1981) - - 0.35




The trunk drainage systems discharging into these basins have been assessed in relation to their performance under these downstream water levels. A desktop review of available construction drawings, reports and DTM of the Study Area has been undertaken. From this a summary has been compiled of estimated maximum allowable downstream water levels in order for:

The �major� system to adequately perform during a 100 year ARI event The �minor� system to adequately perform during a 5 year ARI event

The results of this review are summarised in Table 6-4 below. Table 6-4 Maximum Allowable Pond Flood Levels

Drain Name Wetland System Minor Design Water Level

Minor performance 100 yr Major Design Water

Level

Major performance

South Road Barker Inlet 1.4 (20 yr ARI)

20 year ARI design level exceeded for 5 yr ARI event under existing

conditions.

1.7 Fails under existing conditions. Tolley

Street, Wing Street road levels 1.3mAHD, South Terrace, Leeds Street

1.4mAHD. Dunstan Road Barker Inlet 1.0

(20 yr ARI) Threshold level exceeded

for 5 yr event. 1.7 Threshold exceeded.

Cormack Road and Senna Road most

critical � road levels 1.6-1.7 mAHD.

HEP Barker Inlet 2.15 (20 yr ARI)

Drain is sufficiently high to not be affected by

wetland levels

N/A Drain is sufficiently high to not be affected by

wetland levels North Arm East Barker Inlet 1.9

(20 yr ARI) Drain is sufficiently high to

not be affected by wetland levels

N/A Drain is sufficiently high to not be affected by

wetland levels North Arm Road

The Range 1.35 (upstream

Wilkins Road) (10 yr ARI)

Wetland levels well below maxim allowable levels

1.5 Pond levels below critical catchment

locations. Kapara Road, Wilkins Road road levels 1.5 mAHD.

Hanson Road The Range 0.85 (5 yr ARI)

Pond levels do not exceed outlet pipe obvert for all 5 yr ARI scenarios

1.6 Pond levels well below critical catchment

locations. Hines Road 1.8 mAHD.

Magazine Creek

Magazine Creek 0.10 (10 yr ARI)

Just performs currently, exceeded by future

impacts. Jenkins Street drain design based on 10 yr level of 0.15 mAHD at

Eastern Parade.

0.70 Acceptable performance. Bedford Street is reported to be

the critical location, lowest floor level is 0.83

mAHD.




6.4 Further Work

Further work in the development of a management strategy for the wetland and ponding basin areas needs to take into account:

The challenges and opportunities presented by development proposals, such as developments proposed to be located within the ponding basins

Future increases in flows discharged to these areas due to increased catchment development and climate change

It is clear that proposals to develop in this area will drive the need for the development of a management strategy to ensure that this area can continue to provide an appropriate flood mitigation function taking into account potential future changes to inflows. The model established to determine the existing scenario, as presented in this report, would serve as an excellent tool to support the development of such a strategy. Ongoing development of hydrodynamic models of the stormwater catchments entering the wetlands also present the opportunity to replace the ILSAX hydrology with outflows from these models. This is expected to result in a reduction in flows in some locations and would serve to provide a vastly improved representation of flows entering the wetlands. Further detailed assessment is required of the lower portions of the South Road and Dunstan Road catchments, to confirm critical floor levels in this region and how drainage of this area should be managed given the relatively high Barker Inlet wetland flood levels.

Summary



7. Summary

The aim of the Flood Risk Management Study is to identify the risks and develop and implement a strategy to protect the vulnerable areas of the City from the risk of seawater and stormwater flooding taking into account the possible sea level rise and land subsidence over the next hundred years. Phase 1 of this Study (Risk Assessment and Preliminary Treatment) has been completed. This Phase has analysed and evaluated risk of seawater and stormwater flooding and identified concept management strategies. A land subsidence rate of 2.1 mm/yr has been adopted over the Study Area for this Study. While this over-estimates subsidence in some parts of the Study Area, it is considered appropriate to provide some conservatism (albeit small) particularly in risk assessment. It is recognised that subsidence within the Gillman region is expected to be greater due to local factors. However, uncertainty remains on the current rate (in the range between 2.1 � 10 mm/yr). A statistical analysis was carried out to predict the 100 year storm tide level and duration off-shore of Barker Inlet. The findings from the analysis were used to synthesise a storm tide hydrograph (sea water levels versus time) for application to a hydrodynamic model that can predict over time seawater inundation of low lying areas around Port Adelaide. Floodplain mapping has been undertaken (for a range of future sea level rise scenarios) of inundation from a 100 year ARI tide event. Damage estimates associated with each of these flood scenarios have also been prepared. These predict that the damages associated with a 100 year ARI tide event will increase dramatically from existing conditions ($8m-$28m) to future scenarios associated with 500-880 mm of sea level rise ($180m-$310m). A sea defence upgrade concept design has been developed to protect existing development from this flood risk. The concept design relies on establishing seawalls, of varying type and height. Indicative cost estimates for this concept range from $24m-$31m, depending on the adopted future design scenario. A number of Council drainage systems have been reviewed to identify works required within these catchments to achieve an adequate level of protection against the combined seawater and stormwater flood risk. In several low-lying catchments, stormwater pump stations have been proposed to replace existing gravity stormwater drainage infrastructure. Indicative construction cost estimates for these works are $11m-$18m, depending on the adopted future design scenario.

Summary



Stormwater runoff into the Barker Inlet, Range, and Magazine Creek wetlands, and associated ponding basins was analysed to determine the flood depths and heights for the 100 year event, taking into account seawater interaction. 100 year ARI flood levels in some of the basins are higher than would be preferred for protection to existing development. Future development in this region will need to carefully consider the information contained in this report to ensure that new development is adequately protected and that the flood mitigation performance of the ponding basins is not compromised. The need for further investigations, design development and assessment in a number of areas has been identified. The further development of the seawall concept in particular will require significant effort to progress this Study through Phases 2 (Design) and 3 (Implementation).

References



8. References

Ahmer, I.R. et al (2002), Coastal Flood Modelling: Allowing for Dependence Between Rainfall and Tidal Anomaly, Engineering and Mathematics Applications Conference. Belperio, A.P. (1985), Site investigation in the vicinity of the Dean Rifle Range, Port Adelaide Estuary, Department of Mines and Energy, South Australia, Report 85/54. Belperio, A.P. (1993), Land Subsidence and sea level rise in the Port Adelaide estuary: Implications for monitoring the greenhouse effect, Australian Journal of Earth Sciences, vol 40 p359-368. Belperio, A.P. & Rice R.L. (1989), Cainozoic stratigraphy and structure of the Gawler and Vincent 1:50,000 map sheet areas, Department of Mines and Energy, South Australia, Report 89/62. Culver, R. (1970), Summary Report � Beach Erosion Assessment Study, Department of Civil Engineering - University of Adelaide. Department for Environment and Heritage (2002), Adelaide Coastal Land Level Changes Technical Report (Draft), Coast and Marine Branch, National Parks and Wildlife. Department for Primary Industries and Resources (2001), Southern South Australia Acid Sulphate Potential Mapping, Land Information Unit, South Australia. Engineers Australia (1987), Australian Rainfall and Runoff � Volume 2. Gerges, N. (1996), Overview of the Hydrogeology of the Adelaide Metropolitan Area, Department of Mines and Energy, South Australia. IPCC (2001), "Summary for Policymakers - A Report of Working Group 1 of the Intergovernmental Panel on Climate Change", WMO / UNEP. Kemp, D.J. & Lipp, W.R. (2004) �Predicting Storm Runoff in Adelaide � What Do We Know?�, Transport SA. Postma D. (1983), Pyrite and siderite oxidation in swamp sediments, Journal of Soil Science 34, 163-182.

References



Stephens, J.C. et al (1984), Organic soil subsidence, Reviews in Engineering Geology 6, 107-122. Stelling, G.S. (1984) On the Construction of Computational Methods for Shallow Water Flow Problems Rijkswaterstaat Communications, No 35/1984, The Hague, The Netherlands Syme, W.J. (1991) �Dynamically Linked Two-Dimensional / One-Dimensional Hydrodynamic Modelling Program for Rivers, Estuaries & Coastal Waters� William Syme, M.Eng.Sc (Res) Thesis, Dept of Civil Engineering, The University of Queensland, May 1991. Tonkin Consulting (2003a), Port Road Initial Urban Stormwater Master Plan, City of Port Adelaide Enfield. Tonkin Consulting (2003b), North Arm East Initial Urban Stormwater Master Plan, Cities of Port Adelaide Enfield and Charles Sturt. Tonkin Consulting (2003c), Hart Street Initial Urban Stormwater Master Plan, City of Port Adelaide Enfield. Tonkin Consulting (2003d), TRDA Initial Urban Stormwater Master Plan, Cities of Port Adelaide Enfield and Charles Sturt. Tonkin Consulting (2004), HEP Initial Urban Stormwater Master Plan, Cities of Prospect, Port Adelaide Enfield and Charles Sturt. WBM (2003a), Discussion Paper � Storm Tide Analysis, City of Port Adelaide Enfield WBM (2003b), Discussion Paper 2 � Hydraulic Modelling, City of Port Adelaide Enfield

Appendix A



Appendix A Historical Upper Tertiary (T1) Aquifer Potentiometric Surface Levels (Gerges, 1996)

Appendix B



Appendix B Storm Tide Analysis Additional Information

Appendix B



Appendix C



Appendix C Historical Storm Tide Meteorological Profiles

The following historical storm tide events were short-listed as being suitable for calibration purposes:

18.05.1953; 12.05.1960; 01.06.1981; 03.07.1981; 23.06.1987; 23.05.1988; 13.07.1995; and 13.06.1999.

B144339 Port Adelaide Storm Tide Statistics

G:\Admin\B14339.g.wjs\Extreme Storm Profiles 01 2003-02-19.doc 19/02/2003 RMN

1

APPENDIX: Meteorological Event Profiles

Guide to Synoptic Charts



2

Storm Event 18/05/1953

Outer Harbour Tidal Data - 18/05/1953

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95 0

:00

12/0

7/19

95 6

:00

12/0

7/19

95 1

2:00

12/0

7/19

95 1

8:00

13/0

7/19

95 0

:00

13/0

7/19

95 6

:00

13/0

7/19

95 1

2:00

13/0

7/19

95 1

8:00

14/0

7/19

95 0

:00

14/0

7/19

95 6

:00

14/0

7/19

95 1

2:00

14/0

7/19

95 1

8:00

15/0

7/19

95 0

:00

Date/Time

Win

d Sp

eed

(km

/h)

Mean Sea Level Pressure - 13/07/1995

980

985

990

995

1000

1005

1010

12/0

7/19

95 0

:00

12/0

7/19

95 6

:00

12/0

7/19

95 1

2:00

12/0

7/19

95 1

8:00

13/0

7/19

95 0

:00

13/0

7/19

95 6

:00

13/0

7/19

95 1

2:00

13/0

7/19

95 1

8:00

14/0

7/19

95 0

:00

14/0

7/19

95 6

:00

14/0

7/19

95 1

2:00

14/0

7/19

95 1

8:00

15/0

7/19

95 0

:00

Date/Time

Pres

sure

(hpa

)

Wind Direction - 13/07/1995

0

45

90

135

180

225

270

315

360

12/0

7/19

95 0

:00

12/0

7/19

95 6

:00

12/0

7/19

95 1

2:00

12/0

7/19

95 1

8:00

13/0

7/19

95 0

:00

13/0

7/19

95 6

:00

13/0

7/19

95 1

2:00

13/0

7/19

95 1

8:00

14/0

7/19

95 0

:00

14/0

7/19

95 6

:00

14/0

7/19

95 1

2:00

14/0

7/19

95 1

8:00

15/0

7/19

95 0

:00

Date/Time

Win

d D

irect

ion

(deg

rees

true

)



14






0

2

4

6

8

10

12

14

16

18

20

22

24

26

28

30

32

11/0

6/19

99

12/0

6/19

99

13/0

6/19

99

14/0

6/19

99

15/0

6/19

99

Date/Time

Dai

ly R

ainf

all (

mm

)




15



0

0.5

1

1.5

2

2.5

3

3.5

4

12/0

6/19

99 0

:00

12/0

6/19

99 6

:00

12/0

6/19

99 1

2:00

12/0

6/19

99 1

8:00

13/0

6/19

99 0

:00

13/0

6/19

99 6

:00

13/0

6/19

99 1

2:00

13/0

6/19

99 1

8:00

14/0

6/19

99 0

:00

14/0

6/19

99 6

:00

14/0

6/19

99 1

2:00

14/0

6/19

99 1

8:00

15/0

6/19

99 0

:00

Date/Time

Wat

er L

evel

(m)

Observed Water Level Residual Water LevelPredicted Tide Series4

Average Wind Speed over previous 10 minutes - 13/06/1999

0

10

20

30

40

50

12/0

6/19

99 0

:00

12/0

6/19

99 6

:00

12/0

6/19

99 1

2:00

12/0

6/19

99 1

8:00

13/0

6/19

99 0

:00

13/0

6/19

99 6

:00

13/0

6/19

99 1

2:00

13/0

6/19

99 1

8:00

14/0

6/19

99 0

:00

14/0

6/19

99 6

:00

14/0

6/19

99 1

2:00

14/0

6/19

99 1

8:00

15/0

6/19

99 0

:00

Date/Time

Win

d Sp

eed

(km

/h)

Mean Sea Level Pressure - 13/06/1999

1000

1005

1010

1015

1020

1025

1030

12/0

6/19

99 0

:00

12/0

6/19

99 6

:00

12/0

6/19

99 1

2:00

12/0

6/19

99 1

8:00

13/0

6/19

99 0

:00

13/0

6/19

99 6

:00

13/0

6/19

99 1

2:00

13/0

6/19

99 1

8:00

14/0

6/19

99 0

:00

14/0

6/19

99 6

:00

14/0

6/19

99 1

2:00

14/0

6/19

99 1

8:00

15/0

6/19

99 0

:00

Date/Time

Pres

sure

(hpa

)

Wind Direction - 13/06/1999

0

45

90

135

180

225

270

315

360

12/0

6/19

99 0

:00

12/0

6/19

99 6

:00

12/0

6/19

99 1

2:00

12/0

6/19

99 1

8:00

13/0

6/19

99 0

:00

13/0

6/19

99 6

:00

13/0

6/19

99 1

2:00

13/0

6/19

99 1

8:00

14/0

6/19

99 0

:00

14/0

6/19

99 6

:00

14/0

6/19

99 1

2:00

14/0

6/19

99 1

8:00

15/0

6/19

99 0

:00

Date/Time

Win

d D

irect

ion

(deg

rees

true

)

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