investigation of glass fibre reinforced … · ii investigation of glass fibre reinforced polymer...
TRANSCRIPT
INVESTIGATION OF GLASS FIBRE REINFORCED POLYMER REINFORCING BARS AS INTERNAL REINFORCEMENT FOR
CONCRETE STRUCTURES
by
David Tse Chuen Johnson
A thesis submitted in conformity with the requirements
for the degree of Master’s of Applied Science
Graduate Department of Civil Engineering
University of Toronto
© Copyright by David Tse Chuen Johnson (2009)
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Investigation of Glass Fibre Reinforced Polymer Reinforcing Bars as Internal Reinforcement for Concrete Structures
Master’s of Applied Science
David Johnson
Department of Civil Engineering University of Toronto
2009
ABSTRACT
A study of the existing data shows that two areas of GFRP bar research among others
are in need of investigation, the first being behaviour of GFRP bars at cold temperatures and
the second being the behaviour of large diameter GFRP rods. Based on the results of
experimental work performed, cold temperatures were found to have minimal effect on the
mechanical properties of the GFRP bars tested. In addition, through beam testing, large
32mm diameter GFRP bars were found to not fail prematurely due to interlaminar shear
failure. By evaluating the mechanical and durability properties of GFRP bars and behaviour
of GFRP RC, it can be concluded that GFRP appears to be an adequate alternative
reinforcement for concrete structures. Because of high strength, low stiffness and elastic
behaviour of GFRP bars, issues of significant importance for reinforced concrete are bond
development, influence of shear on member behaviour and member deformability.
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ACKNOWLEDGEMENTS
There are many people involved in this research project that without their help, this
project would not be what it is today. First and foremost, I would like to thank Professor
Shamim A. Sheikh for his patience and guidance during this research project and the writing
of this thesis. I would also like to thank the second reader of this thesis Professor Frank J.
Vecchio for his help and constructive comments.
The experimental program for the research project required the assistance of two
separate labs and numerous laboratory support staff. First and foremost, special thanks are
due to the laboratory staff at the University of Toronto Structural Research Labs (Renzo
Basset, Joel Babbin, Giovanni Buzzeo, Alan McClenaghan and John MacDonald). The
assistance also provided by the technical staff of the concrete technology group at Kinectrics
Inc. is also greatly appreciated (Ron Cullen, Joe Aloisio).
The support from engineers and material provided by Schock Bauteile GmbH and
Schock Canada Inc. was critical to the success of the program. I would especially like to
thank Christian Witt, Benjamin Jütte for their continued support during the duration of the
research program. In addition, special thanks are due to Dr. André Weber of Schock
Bauteile GmbH for his support and for providing the research reports used in this thesis.
Financial support from the University of Toronto, Government of Ontario, NSERC Canada
and the ISIS Research Network was greatly appreciated.
Finally, to all of my colleagues and friends in the Department of Civil Engineering,
your support and friendship was invaluable. Present and former members of the FRP
research group at U of T (Michael Colalillo, Alex Caspary, Jingtao Liu, Sylvio Tam and
Ciyan Cui) helped through all stages of the program whether it be the casting of specimens or
just being there to bounce ideas off of, for that you were invaluable and I thank you all. The
work of undergraduate research students Junghyun Park and Arjang Tavassoli was also
greatly appreciated. As well as fellow research students at the University of Toronto who
also helped with the casting and testing of the specimens namely Boyan Mihailov and Jimmy
Susetyo.
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I would also like to especially thank Karen Woo for not only her help in the lab but
also her continued support during the entire degree.
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TABLE OF CONTENTS 1 OBJECTIVE AND SCOPE ................................................................................................. 1
1.1 Research Signficance ................................................................................................. 1
1.2 Corrosion Case Study (America’s Bridges) ............................................................... 2
1.2.1 The extent of corrosion in bridges ...................................................................... 2
1.2.2 Recent progress on corrosion mitigation ............................................................ 3
1.2.3 Extent of corrosion damage in Canada ............................................................... 4
1.2.4 GFRP as a potential solution .............................................................................. 5
1.3 Scope of the Research Program ................................................................................. 6
2 BACKGROUND ON GFRP REINFORCING BARS .................................................... 7
2.1 Fibre Materials ........................................................................................................... 7
2.2 Resin Materials .......................................................................................................... 9
2.3 Fabrication Techniques .............................................................................................. 9
2.4 ISIS Certification Standard and CSA S807 ............................................................... 9
2.5 Mechanical Properties of Glass Reinforcing Bars. .................................................. 11
2.5.1 Available Bars .................................................................................................. 12
2.5.2 Glass Transition Temperature (Tg) .................................................................. 14
2.5.3 Cure Ratio ......................................................................................................... 15
2.5.4 Other reinforcement products ........................................................................... 18
2.5.5 Summary ........................................................................................................... 18
3 LITERATURE REVIEW .................................................................................................... 19
3.1 Previous Work on the Flexural Behaviour ............................................................... 19
3.1.1 Nawy et al 1971, 1977 ...................................................................................... 21
3.1.2 Brown and Bartholomew 1993 ......................................................................... 21
3.1.3 Benmokrane, Challal and Masmoudi 1996 ...................................................... 22
3.1.4 Vijay and Gangarao 2001 ................................................................................. 22
3.1.5 Yost, Gross and Dinehart 2003 ......................................................................... 23
3.1.6 General conclusions on the flexural behaviour ................................................ 24
3.2 Previous Work on Bond ........................................................................................... 24
3.2.1 Malvar 1995 ...................................................................................................... 25
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3.2.2 Tastani and Pantazopoulou 2002 ...................................................................... 25
3.2.3 Achillides and Pilakoutas 2004 ........................................................................ 26
3.2.4 Wambeke and Shield 2006 ............................................................................... 27
3.2.5 Mosley, Tureyan and Frosch 2008 ................................................................... 27
3.2.6 General conclusions on the bond behaviour ..................................................... 28
3.2.7 Summary ........................................................................................................... 28
4 DURABILITY OF GFRP REINFORCEMENT .............................................................. 30
4.1 Alkali Resistance of GFRP Reinforcing Rods ......................................................... 30
4.1.1 Alkali resistance and testing ............................................................................. 30
4.1.2 Alkali resistance of commercially available bars ............................................. 31
4.2 Creep Rupture Strengths .......................................................................................... 33
4.2.1 Creep rupture test method (CSA S806-02) ....................................................... 33
4.2.2 Creep rupture strength of Available GFRP bars ............................................... 34
4.3 Performance in Extreme Temperature Environments .............................................. 37
4.3.1 Glass Transition Temperature (Tg) .................................................................. 37
4.3.2 Bar mechanical property change under extreme heat ....................................... 38
4.3.3 Bond strength degradation under extreme heat ................................................ 40
4.3.4 Response of GFRP Reinforcing Bars to Extreme Cold .................................... 42
4.4 Fatigue Strength of GFRP Reinforcing Rods .......................................................... 42
4.4.1 Test method and results of fatigue testing ........................................................ 42
4.5 Do Simulated Lab Tests Reflect the True Conditions? ............................................ 45
4.6 Summary of Durability ............................................................................................ 45
5 EXPERIMENTAL WORK ................................................................................................. 47
5.1 GFRP Extreme Cold Temperature Tests ................................................................. 47
5.1.1 Objective of cold temperature tests .................................................................. 47
5.1.2 Specimen preparation ....................................................................................... 47
5.1.3 Control sample testing at room temperature ..................................................... 49
5.1.4 Cold temperature test setup .............................................................................. 51
5.1.5 Specimen mounting .......................................................................................... 52
5.1.6 Instrumentation ................................................................................................. 53
5.1.7 Testing procedure ............................................................................................. 54
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5.2 GFRP-Reinforced Large Beams .............................................................................. 54
5.2.1 Objective of the test .......................................................................................... 54
5.2.2 Specimen design ............................................................................................... 54
5.2.3 Construction and casting of the beams ............................................................. 55
5.2.4 Instrumentation ................................................................................................. 57
5.2.5 Test setup and procedure .................................................................................. 58
6 RESULTS AND DISCUSSION ....................................................................................... 61
6.1 Cold Temperature Test Results ................................................................................ 61
6.1.1 Results of the tensile tests on 8mm bars ........................................................... 61
6.1.2 Results of the tensile test on 12mm bars .......................................................... 64
6.1.3 Results of tensile tests on 16mm ...................................................................... 65
6.2 Large Beam Tests..................................................................................................... 68
6.2.1 Load deflection response .................................................................................. 68
6.2.1 Moment-curvature response ............................................................................. 70
6.2.2 Bar stress-strain response ................................................................................. 71
6.2.3 Crack width behaviour ...................................................................................... 77
6.2.4 Bond and stress development behaviour .......................................................... 77
6.3 Prediction of Large Beam Samples .......................................................................... 84
6.3.1 Sectional analysis (Response 2000).................................................................. 84
6.3.2 Non linear finite element analysis (VecTor2) .................................................. 86
6.3.3 Results of analysis procedures .......................................................................... 87
7 DESIGNING WITH GFRP .............................................................................................. 92
7.1 Canadian Design Codes for GFRP RC .................................................................... 92
7.2 International Codes for Design ................................................................................ 93
7.3 Proposed Design Methodology for GFRP Bars ....................................................... 94
7.3.1 Flexural design ................................................................................................. 95
7.3.2 Designing for Shear with GFRP ....................................................................... 99
7.3.3 Quantifying ductility ......................................................................................... 99
7.3.4 Vijay and Gangarao 2001 (DF Factor) ........................................................... 100
7.3.5 Yost & Gross 2002 (EFS Design) factor and method .................................... 101
7.3.6 CHBDC code J-Factor and design equations ................................................. 102
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7.4 Design Example of a One-way Slab Reinforced with GFRP Bars ........................ 103
7.4.1 Brief Summary of Design ............................................................................... 104
7.4.2 Analysis and Discussion of Sample Design ................................................... 105
7.4.3 Moment-Shear Interaction .............................................................................. 108
7.5 Hybrid Section Design ........................................................................................... 110
7.5.1 Principle of hybrid design ............................................................................... 110
7.5.2 Comparative study of reinforcement types and layouts ................................. 110
7.6 Summary on the Design of GFRP-Reinforced Concrete Members ....................... 114
8 CONCLUSIONS .............................................................................................................. 115
8.1 General Conclusions on GFRP bars and GFRP Reinforced Concrete ................... 115
8.2 Future work ............................................................................................................ 116
9 REFERENCES ................................................................................................................... 118
Test and Technical Reports .......................................................................................... 118
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LIST OF TABLES
Table 1-1 Summary of Deficiency in the American Bridge Inventory (FHWA, 2006) .................. 3
Table 1-2 Structurally Deficient Bridges per state for those bordering Canada .............................. 4
Table 2-1 Types of Available Structural Glass Fibres ..................................................................... 8
Table 2-2 Chemical Composition and Properties of Various Glass Types. (Ehrenstein 2007) ..... 8
Table 2-3 Minimum Modulus of Elasticity requirements for ISIS Compliance ........................... 10
Table 2-4 Durability Grades and Requirements for Glass FRPs ................................................... 11
Table 2-5 Mechanical Properties of ASLAN 100 Reinforcing Bar .............................................. 12
Table 2-6 Mechanical Properties of Pultrall V-Rod Reinforcing Bar ........................................... 13
Table 2-7 Mechanical Properties of Pultrall V-ROD HM Reinforcing Bar .................................. 13
Table 2-8 Mechanical Properties of Schöck ComBAR Reinforcing Bars .................................... 14
Table 2-9 Glass Transition Temperatures for Schöck ComBAR .................................................. 15
Table 3-1 Selected Flexural Behaviour Tests on GFRP RC.......................................................... 20
Table 4-1 Summarized Alkali Resistance of Commercially Available GFRP bars. ..................... 32
Table 4-2 Summarized Creep Rupture Data for 16mm ComBAR bars ........................................ 34
Table 4-3 Linear Regression of Creep Rupture Data for 16 mm Bar ............................................ 35
Table 4-4 Predicted Millionth Hour Creep Strengths for 16mm Bar ............................................ 36
Table 4-5 Creep Rupture Data for 8mm and 25mm ComBAR bars ............................................. 36
Table 4-6 Summary of Glass Transition Temperature Results ..................................................... 38
Table 4-7 Cyclic Bending Tests Results ........................................................................................ 44
Table 5-1 Samples for Cold Temperature Testing ........................................................................ 48
Table 6-1 Results of 8mm Cold Temperature Tests ...................................................................... 62
Table 6-2 Summary of results for 8mm tests ................................................................................ 62
Table 6-3 Test Results for 12mm ComBAR Samples ................................................................... 64
Table 6-4 Summarized Results of 12mm Tests ............................................................................. 65
Table 6-5 Test Results for 16mm ComBAR Samples ................................................................... 66
Table 6-6 Summarized Results of Select 16mm Tests .................................................................. 67
Table 6-7 Cracking Load, Moment and Midspan Displacement for all samples .......................... 69
Table 6-8 Failure Load, Moment and Midspan Displacement for all samples ............................. 70
Table 6-9 Peak Bar Stresses for all Large Beam Tests .................................................................. 72
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Table 6-10 Estimates of GFRP Bar Modulus of Elasticity for 5 Large beams ............................. 75
Table 6-11 Summary of modulus of elasticity estimates............................................................... 77
Table 6-12 Summary of Calculated Bond Strengths ..................................................................... 81
Table 6-13 Pull-Out Bond Strengths for 16mm GFRP Bar ........................................................... 82
Table 6-14 Summary of parameters in Response 2000 analysis ................................................... 85
Table 7-1 Key differences between S806-02 and S6-06 ............................................................... 92
Table 7-2 Materials Resistance Factors and Stress Limits from Various Codes for GFRP
Design ............................................................................................................................................ 94
Table 7-3 Key Design Details for Sample Slab ........................................................................... 104
Table 7-4 Summary of performance measures for sample slab .................................................. 108
Table 7-5 Key for section names ................................................................................................. 111
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LIST OF FIGURES
Figure 2-1 Pultrall V-Rod, Hughes Bros. Aslan 100 FRP and Schöck ComBAR (Left to
Right) ...................................................................................................................................... 11
Figure 2-2 Photograph of V-Rod and V-Rod HM (Courtesy B. Benmokrane)...................... 14
Figure 2-3 Tensile Strength vs. Cure Ratio for configurations of resin and laminates (Hülder,
2008) ....................................................................................................................................... 17
Figure 3-1 Idealized normal stress distribution (Achillides, 2004) ........................................ 26
Figure 4-1 Setup for Creep Rupture Tests on ComBAR bars (Weber, 2005) ........................ 33
Figure 4-2 Plot of Sustained Stress vs Failure Time in Creep Rupture Tests ........................ 35
Figure 4-3 Comparison of ComBAR sizes in creep rupture................................................... 37
Figure 4-4 Tensile Strength vs. Temperature for ComBAR bars (Nause 2005) .................... 39
Figure 4-5 Tensile Strength vs. Temperature for V-Rod bars (Robert et al. 2009) ................ 40
Figure 4-6 Pull-out and Push-through bond testing at various temperatures (Weber 2008) .. 41
Figure 4-7 Load Slip Charts for ASLAN 100 FRP under varying temperature conditions.
(Katz et al. 1999) .................................................................................................................... 41
Figure 4-8 Specimens and Reinforcing for Dynamic Tests at Karlsruhe (Kreuser 2007) ..... 43
Figure 4-9 Test Setups for fatigue strength testing at Karlsruhe (Kreuser 2007) ................... 43
Figure 5-1 Schöck ComBAR specimens with attached couplers ........................................... 48
Figure 5-2 Thermotron Unit used for preconditioning ........................................................... 49
Figure 5-3 Overview of Control Sample test setup (16mm sample shown (TCB16-01)) ...... 50
Figure 5-4 Universal Test Machine with Attached Environmental Chamber ........................ 51
Figure 5-5 BEMCO Control Unit and Thermocouple Readout.............................................. 52
Figure 5-6 ComBAR Sample Mounted into Environmental Chamber (TCB16-03 shown) .. 53
Figure 5-7 Geometry of Large Beam Samples ....................................................................... 55
Figure 5-8 Formwork for the large beams .............................................................................. 56
Figure 5-9 GFRP wrapped beams .......................................................................................... 57
Figure 5-10 Test setup for large beam tests ............................................................................ 58
Figure 5-11 MTS Machine used for testing ............................................................................ 58
Figure 5-12 Load arrangements for large beam tests ............................................................. 59
Figure 5-13 Picture of TCB 3202 before application of load. ................................................ 60
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Figure 5-14 Picture of TCB 3202 during testing. ................................................................... 60
Figure 6-1 Typical debonding failure ..................................................................................... 63
Figure 6-2 Typical rupture failure .......................................................................................... 63
Figure 6-3 Photograph of ruptured coupler in sample TCB12-04 .......................................... 65
Figure 6-4 Load Deformation Response of All Large Beam Samples ................................... 68
Figure 6-5 Moment curvature responses for all 5 beams ....................................................... 71
Figure 6-6 GFRP bar stress strain plots for TCB3201 & 3202 .............................................. 73
Figure 6-7 GFRP bar stress strain plots for TCB3203 ........................................................... 74
Figure 6-8 GFRP bar stress strain plots for TCB3204 & 3205 .............................................. 74
Figure 6-9 Bar Stress and Moment Diagram for TCB3201 ................................................... 78
Figure 6-10 Bar Stress and Moment Diagram for TCB3202 ................................................. 79
Figure 6-11 Bar Stress and Moment Diagram for TCB3203 ................................................. 79
Figure 6-12 Bar Stress and Moment Diagram for TCB3204 ................................................. 80
Figure 6-13 Bar Stress and Moment Diagram for TCB3205 ................................................. 80
Figure 6-14 Debonded GFRP bar inside large beam post failure ........................................... 84
Figure 6-15 Response 2000 moment curvature prediction with experimental results ........... 85
Figure 6-16 Mesh for VecTor 2 Analysis ............................................................................... 87
Figure 6-17 Load deflection of TCB3201 & 3202 with software analysis predictions.......... 88
Figure 6-18 Load deflection of TCB3203 with software analysis predictions ....................... 89
Figure 6-19 Load deflection of TCB3204 & 3205 with software analysis predictions.......... 90
Figure 7-1 Flow Chart for Tensile Rupture Controlled Design Flexural Strength Calculation
................................................................................................................................................ 98
Figure 7-2 Geometry and Loading of Sample Slab .............................................................. 104
Figure 7-3 Moment Curvature Response for Sample Slab ................................................... 106
Figure 7-4 Areas for determination of energy dissipation .................................................... 107
Figure 7-5 Moment Shear Interaction for Sample Slab ........................................................ 109
Figure 7-6 Concrete Section used for Hybrid Section Analysis ........................................... 111
Figure 7-7 Moment Curvature Responses for all 4 Sections ................................................ 112
Figure 7-8 Enlarged Low Curvature Region of Moment Curvature Responses .................. 113
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LIST OF APPENDICIES
Appendix A – Stress-Strain plots for Cold Temperature Samples
Appendix B – Beam Photos
Appendix C – Model Parameters for VecTor 2 Analysis
Appendix D – Sample Design of Rupture Controlled GFRP RC Slab
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1 OBJECTIVE AND SCOPE
1.1 Research Signficance
Infrastructure in Canada is aging and deteriorating; the costs of repair and
rehabilitation are a constant strain on the already limited available public funds. The
replacement cost of Ontario’s bridges and highways is estimated to be approximately 57
billion (MTO, 2009). Canada spends a significant amount of money annually on its
infrastructure but even at that pace cannot clear up the backlog of work that needs to be done
on not only its bridges, but infrastructure in general. Even in poor economic times public
infrastructure remains a top spending priority, and that money has to be well spent and used
as effectively as possible.
Aging infrastructure and the inflating costs to maintain them is not only a Canadian
problem, it’s a worldwide issue. In a recent study in 2006 from the federation international
du beton (fib), their task force 9.3 on reinforced concrete estimated that worldwide
infrastructure maintenance and repair exceeds 100 billion euros ($155 billion CDN) annually.
Consider the United States of America, the Federal Highway Administration (FHWA) in
their recent 2007 bridge inventory check identified 594,806 total bridges in America, of
which 79,635 (13.4%) were functionally obsolete and 72,178 (12.05%) were deemed
structurally deficient, equating to a staggering total of 151,813 (25.5%) deficient bridges.
The American Society of Civil Engineers (ASCE) in their report card on American
Infrastructure in 2003 indicated that a total investment of 1.3 trillion US dollars would be
required over 5 years to bring infrastructure in America up to an acceptable level (ASCE,
2003). While Canada does not have nearly as accurate an idea of its inventory, the trends and
problems are very similar and needs to be fixed.
Reinforced concrete since its inception in 1893 has been plagued by the problems
associated with the corrosion of the steel reinforcing bars. With advances in polymer
technology, Glass Fibre Reinforced Polymer (GFRP) bars are emerging as the viable solution
to the problem of steel corrosion in newly built and rehabilitated structures. With all the
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delays and deferrals of rehabilitation and inspection work, Canada can’t afford to continue
building structures that are still vulnerable to corrosion. The 2006 collapse of the Laval
overpass and even more recently the Saint-Laurent parking garage in Montreal in 2008 are
both sobering reminders of how important it is to inspect and rehabilitate our aging
infrastructure effectively both cost and performance wise.
Stainless steel reinforcing, while effective at mitigating corrosion, is proving too costly
a material for widespread use in all structures. Reinforced concrete has evolved over the past
century with such advances as air entrainment in concrete, epoxy coated reinforcing and the
most recent use of stainless steel reinforcing. GFRP is the next advancement in that line, with
new certification standards available and manufacturers producing bars of consistent quality,
it is essential that the engineering community understand this new material as much as
possible to facilitate its integration into Canadian and international infrastructure and derive
benefits from its durable performance.
1.2 Corrosion Case Study (America’s Bridges)
1.2.1 The extent of corrosion in bridges
While it would be preferable to consider Canadian bridges in the case study, however
having no national inventory or national body makes a country wide study impossible. The
U.S. Federal Highway Administration (FHWA 2007) has identified in its reports over the
past decade that corrosion is one of the largest contributors to structural deficiency among
the bridges in the United States national bridge inventory. A structural deficiency as defined
by the FHWA is characterized by “deteriorated conditions of significant bridge elements and
reduced load carrying capacity” (FHWA 2007). Corrosion can be assumed to affect any and
every concrete bridge to some extent, however, in this section only bridges that have
deteriorated to a condition that qualifies as structurally deficient will be considered.
By analysing the bridge inventory it can be seen that 64% or 380,772 bridges have
their primary load carrying components made out of concrete (either pre-stressed or
reinforced). Of those 380,772 bridges only 23,542 (6%) are structurally deficient. That
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number is much better compared to the overall 12.05% of all bridges that are structurally
deficient. Structural deficiency of a key component like a concrete deck slab on a steel bridge
can also qualify the bridge to be structurally deficient. Steel bridges comprise 139,445
(23.27%) of the total inventory of which a very high 34,057 (23.5%) are structurally
deficient.
1.2.2 Recent progress on corrosion mitigation
The large number of deficient bridges is one major issue; more importantly however, is
the annual backlog of bridges that need maintenance. The FHWA over the past 20 years has
been annually keeping track of how many deficient bridges exist in their system. Shown in
the Table 1-1 is a snapshot of them over the past few years taken from the 2006 FHWA
report to congress showing the rates of deficiency by bridge class during a recent 10 year
period.
Table 1-1 Summary of Deficiency in the American Bridge Inventory (FHWA, 2007)
Year 1994 1996 1998 2000 2002 2004
Interstate Roads
SD 6.0 6.0 5.4 5.2 5.1 5.1 FO 18.2 18.7 16.2 16.4 16.0 16.1
Arterial Roads
SD 10.9 10.2 9.3 8.3 8.0 7.7 FO 10.7 10.9 10.8 11.0 11.0 10.7
Collector Roads
SD 16.1 14.9 14.0 13.2 12.5 12.0 FO 11.7 11.4 11.8 11.7 11.9 11.6
Local Roads
SD 27.9 25.9 23.5 21.9 20 19 FO 12.4 12.1 12.5 12.5 12.7 12.3
SD – Structurally Deficient, FO – Functionally Obsolete, *All numbers are percentages
What is clear in the above table is that there is a backlog of bridge work that needs to be
done. While progress is being made in some areas, most notably local and collector roads,
the rates of deficiency are still quite high.
One alarming trend that can be quickly found by scanning through the bridge inventory
is that many structurally deficient bridges were built between 1993 and 1997. Bridges as
little as 10 years old are already becoming structurally deficient. In the case of two states:
4
Mississippi and Oklahoma, there are 178 and 106 currently deficient bridges built in that 4
year span, respectively. The backlog of deficient bridges cannot be completely dealt with if
structures are deteriorating significantly in 10 years time after construction.
1.2.3 Extent of corrosion damage in Canada
Canada is well known for its significantly colder winters and liberal use of deicing salts,
whose detrimental effects on reinforced concrete structures are well researched and
documented. It is unclear at this time just how many bridges are being affected because of a
lack of book keeping at the government level and a lack of a central co-ordinating and
governing body.
Comparing the states that border Canada in the FHWA inventory can provide some
insight into the effects de-icing salts and colder climates have on their bridges. Listed in
Table 1-2 are the percent of structurally deficient bridges in cold climate states that border
Canada as of December 2007.
Table 1-2 Structurally Deficient Bridges per state for those bordering Canada
State # of Bridges and % of State Inventory that is Structurally Deficient
Michigan 1, 569 (14.39 %)
Minnesota 1,149 (8.8 %)
Montana 472 (9.48 %)
New Hampshire 380 (16.12 %)
New York 2,118 (12.23 %)
North Dakota 741 (16.64 %)
Illinois 2,482 (9.56 %)
Ohio 2,842 (10.17 %)
Pennsylvania 5,757(25.9 %)
Vermont 500 (18.5 %)
5
With the exception of 4 states, all of the states in the above list are above the national
average. With consistent standards and rating according to the National Bridge Inspection
Standards, it can be concluded that the presence of de-icing salts and extreme conditions do
indeed increase the rates of structural deficiency in bridges to some extent depending on the
quality of the maintainers. One can expect that in Canada the rates would be similar if not
higher because of a longer winter.
1.2.4 GFRP as a potential solution
Stainless steel bars are being introduced into newly constructed reinforced concrete
bridges to help stop the corrosion of newly built structures. Stainless steel is significantly
more expensive than regular steel and it is not currently feasible to use it in all the bridges
that need to be built annually. The cost of stainless steel reinforcing bars can be estimated to
be 5 to 6 times greater than a traditional carbon steel bar (Russell, 2004) which can roughly
translate into an additional 10-15% of the initial capital costs of the bridge. The costs of
GFRP on the other hand, are competitive depending on the manufacturer. Research
conducted by the National Composites Network in Europe (Halliwell, 2002) has shown that
GFRP reinforcing bars cost about half of what stainless steel costs. The cost of GFRP bars in
recent years has been coming down primarily due to a larger market and competition.
The integration of GFRP into infrastructure has generally been delayed because of a
lack of any set standards for manufacturing or design. That problem will change with the
new Canadian Standard CSA S807 which complements the current RILEM and ACI
standards in Europe and America. With a competitive marketplace, standards in place, pilot
project bridges, and significant research and development: GFRP is emerging as a legitimate
alternative to steel reinforcing. It is important then to fully evaluate GFRP in the context of a
primary reinforcing product, by looking at the short and long-term structural and durability
performance to determine its adequacy as internal reinforcement for concrete structures.
6
1.3 Scope of the Research Program
The overall goal of the study herein is to evaluate the latest generation of GFRP
reinforcing products, as well as identify and evaluate the overall adequacy of using these
products as not only secondary but primary load carrying tensile reinforcement in reinforced
concrete. Only GFRP in typical reinforced concrete applications will be considered,
prestressed as well as complex disturbed region design will not be discussed in this study.
The research program is focused around three primary goals: i) an evaluation of
current generation GFRP products and the current certification guidelines, ii) the testing and
analysing of the structural behaviour of large scale GFRP reinforced members, iii) the
analysis of the current design provisions and design methodologies for GFRP reinforced
members.
In addition, the code provisions in Canada, the United States and Europe for
designing with GFRP will be evaluated. Based on an evaluation of previous research and
experimental results, modifications and refinements to Canadian code provisions will be
proposed. Sound design methodologies for the design of both GFRP and GFRP-steel hybrid
sections will also be introduced.
Finally at the conclusion of the study, the overall suitability of GFRP as concrete
reinforcement will be discussed.
7
2 BACKGROUND ON GFRP REINFORCING BARS
FRP’s of various types and configurations have existed since the post world war II
era (Tang, 1997). Their high strength to weight ratio have made them an attractive building
and structural rehabilitation material. With recent advances in polymer technology, FRP
reinforcing bars can now allow engineers to utilize the benefits of using FRP as internal
reinforcing for concrete structures. Currently there are three major manufacturers of GFRP
reinforcing bars, two North American and one European.
2.1 Fibre Materials
Both FRP wrap systems and FRP reinforcing bars can be made from one of three
typical materials. The most commonly known fibre material of the three is carbon fibre,
famous for its use in other industries including most notably aircraft, formula race car
construction and sporting goods. The other two materials are aramid (Kevlar) and glass. Each
of the three materials has different mechanical and structural properties, which should be
taken into consideration when choosing which material would best suit the application. Some
studies have indicated that the type of material can also influence the resistance to
environmental exposure and in turn the durability. Tam and Sheikh (2007) tested the
durability of various FRP materials to determine their resistance to environmental exposures.
Aramid and carbon FRP reinforcing bars are seldom considered for use in reinforced
concrete because of their significantly higher costs than standard glass. FRP reinforcing bars
made of carbon or aramid will not be dealt with in any detail in this study. As well, it should
be noted that bars made of basalt fibres are also beginning to emerge onto the marketplace.
Within the glass category there are further subdivisions. Shown in Table 2-1 are the
types of glass currently available which are usable in FRPs. The type of glass commonly
used for reinforcing rods is the E-glass type, the exact same type that is used in FRP wrap
systems. E-glass is the material of choice because of its low cost relative to the other
8
available types. A comparison of the chemical compositions and mechanical properties for
all the types of glass are shown below in Table 2-2 (Ehrenstein 2007).
Table 2-1 Types of Available Structural Glass Fibres
Glass Designation Type E-Glass Standard conventional glass type S-Glass High strength glass C-Glass Chemical resistant glass ECR Glass Chemically resistant conventional glass AR-Glass Alkali resistant glass Table 2-2 Chemical Composition and Properties of Various Glass Types. (Ehrenstein 2007)
E-Glass S-Glass C-Glass ECR-Glass AR-Glass Component %
SiO2 54 60 60-65 54-62 62 Al2O3 14-15 25 2-6 12-13 - CaO 20-24 14 14 21 5-9 MgO - 3 1-3 4.5 1-4 B2O3 6-9 <1 2-7 <0.1 <0.5 K2O <1 <1 8 0.6 - Na2O - - - - 12-15 ZrO2 - - - - 17
Properties Density (g/cm3) 2.6 2.53 2.52 2.72 2.68
Tensile Strength (MPa) 3400 4400 2400 3440 3000 Modulus of Elasticity
(MPa) 73000 86000 70000 73000 73000
Ultimate Strain (%) <4.8 <4.6 <4.8 <4.8 <4.4 Thermal Coefficient
(x10-6/oC) 5.0 4.0 6.3 5.9 6.5
Softening Temperature (oC)
850 980 750 880 770
As shown in Table 2-2, the predominant molecular unit in all the glass fibres is
silicon. Silicon is what gives the glass fibres their strength but at the same time can be their
weakness as they are part of an important chemical reaction involving basic hydroxyl ions.
The reaction deteriorates the fibre matrix which in turn may degrade the internal structure of
the bars significantly. Resistance to alkalis will be covered in more detail in section 3.1.
9
2.2 Resin Materials
Resin materials have been shown to significantly influence the mechanical properties
of the overall reinforcing bar. The primary function of the resin is to provide a mechanism of
load transfer into the fibres (Ehrenstein 2007). While older generation products used
polyester resins and epoxy based derivatives of Bisphenol-A (BPA) resins, newer resins are
currently being used.
Today, three major categories of thermosetting resin exist, polyester, epoxy and
vinyl-ester. Typically, polyester resins are thought to be easy to cure but have the lowest
mechanical properties and chemical resistance. On the other hand, epoxy resins are known to
have excellent durability and chemical resistance but require high temperatures to cure
properly. Vinyl-ester (VE) resins have a combination of properties from both epoxy and
polyester resins which make them ideal for using in GFRP bars (fib, 2006).
One modification to the VE resin that has been done is the addition of urethane (VEU)
which provides additional stability in the matrix. VE and VEU resins have been shown to
have improved chemical resistance and enhanced mechanical characteristics. Some evidence
has also shown that vinyl-ester based resins are optimal for GFRP bar manufacturing because
of their easily controllable curing reactions.
2.3 Fabrication Techniques
The primary method of fabrication of GFRP bars is pultrusion. Pultrusion is a very cost
effective method of producing the FRP reinforcing rods initially patented in the early 1950s.
Glass fibre strands are pulled through a resin bath and then heat cured in a steel die. After
cooling, the bars are cut to the desired shape. This method is consistent for all FRP and resin
types.
2.4 ISIS Certification Standard and CSA S807
With the rapid evolution of GFRP reinforcing bars over recent years and the
emergence of new products into the marketplace, certification standards have become
10
necessary. The ISIS (Intelligent Sensing for Innovative Structures) Canada research network
previously developed guidelines for certification for FRP reinforcing bars. The document,
“Specifications for Product Certification of FRPs as Internal Reinforcing for Concrete
Structures” is currently in the process of being modified and adopted by the Canadian
Standards Association (CSA) as standard S807. Most of the provisions found in the ISIS
document are very similar to the American standard ACI 440.3R as both incorporate testing
requirements from ASTM standards (American Society for the Testing of Materials).
FRP bars can be designated at varying grades and quality designations as outlined in
the ISIS and CSA guidelines. Varying strength and stiffness grades exist, which depend on
the results of the tensile tests. A durability designation also exists, which depends on the
results of the specified durability testing. Tables 2-3 and 2-4 below show the different grades
and designations and their requirements.
If adequate, the bars are given a grade and designation which is of the following form:
Ga-Eb-Dc. The G designates that the FRP is of the Glass type, and the letters a, b and c refer
to the different properties. The letter “a” denotes the strength of the bars in MPa, “b” is one
of the three stiffness grades outlined in Table 2-3, while the letter “c” is one of the two
durability grades shown in Table 2-4. ISIS prescribes additional tests that do not have a
bearing on any of the above mentioned designations but are required to meet the overall
certification requirements in general.
Table 2-3 Minimum Modulus of Elasticity requirements for ISIS Compliance
Grade I Grade II Grade III
Minimum modulus of elasticity
requirement (MPa) 50,000 40,000 35,000
11
Table 2-4 Durability Grades and Requirements for Glass FRPs
Durability Grade 1 (D1) Durability Grade 2 (D2)
Maximum Water Absorption (%) 0.75 1.0
Minimum Cure Ratio (%) 98 95
Minimum Glass Transition
Temperature (oC) (DMA) 110, (DSC) 100 (DMA) 90, (DSC) 80
Minimum Alkali Resistance without
Loading (% of Tensile Capacity) 80 70
Minimum Alkali Resistance with
Loading
(% of Tensile Capacity)
70 60
2.5 Mechanical Properties of Glass Reinforcing Bars.
GFRP reinforcing bars are well known for their high strength to weight ratio and linear
elastic stress strain response. While all bars exhibit those characteristics, significant
differences exist from one manufacturer to another. The following sections discuss the
different products currently available and their mechanical properties determined from
previous testing. Photos of all three products are shown in Figure 2-1 below.
Figure 2-1 Pultrall V-Rod, Hughes Bros. Aslan 100 FRP and Schöck ComBAR
(Left to Right)
12
2.5.1 Available Bars
Aslan 100 FRP is a product manufactured by Hughes Brothers Inc., located in
Nebraska, USA. The company has been in business manufacturing timber, steel and
fibreglass products since 1921. The Aslan 100 bar is an E-Glass bar, helically wrapped and
sand coated for enhanced bonding characteristics. Shown in Table 2-5 are the mechanical
properties that were determined from Hughes Brothers own internal testing (Colberg, 2007).
Table 2-5 Mechanical Properties of ASLAN 100 Reinforcing Bar
Nominal Diameter (mm) 13 16 19 22
Cross Sectional Area (mm2) 133 201 284 380
Tensile Strength (MPa) 867 743 749 707
Modulus of Elasticity (MPa) 46100 44600 42500 44000
Ultimate Elongation (%) 1.88 1.67 1.77 1.61
While the modulus of elasticity and the ultimate elongation remained relatively
constant over all the diameters tested, the ultimate strength is generally lower for larger bars.
One of the main reasons for the inverse relationship of strength and bar size is that the fibre
glass is heavily sensitive to small defects and smaller diameters have a smaller chance of
containing or precipitating these defects (Nawy et al. 1971).
V-Rod is manufactured by Pultrall Inc. based in Thetford Mines Quebec. Pultrall
currently has two GFRP reinforcing products, the standard V-Rod bar and the new higher
strength V-Rod HM. The original V-Rod is a sand-coated bar using standard E-Glass fibres.
The mechanical properties of the V-Rod are shown in Table 2-6 below.
13
Table 2-6 Mechanical Properties of Pultrall V-Rod Reinforcing Bar (Pultrall, 2007) Nominal Diameter (mm) 6 10 13 16 19 22 25
Cross Sectional Area (mm2) 28.3 78.5 133 201 284 380 491
Tensile Strength (MPa) 874 856 786 751 728 693 675
Modulus of Elasticity (MPa) 46100 45400 46300 48200 47600 46400 51900
Ultimate Elongation (%) 1.90 1.89 1.70 1.56 1.53 1.49 1.30
Again the strength decreases with the increase of bar size and the properties of this
product are relatively similar to those of the ASLAN 100 bars. The properties of the V-Rod
HM are significantly higher than that standard V-Rod and are shown in Table 2-7 below.
Table 2-7 Mechanical Properties of Pultrall V-ROD HM Reinforcing Bar. (El-Gamal et al. 2008)
Nominal Diameter (mm) 12.7 16 25.4 31.8
Cross Sectional Area (mm2) 127 198 507 791
Tensile Strength (MPa) 1450 1439 1260 1060
Modulus of Elasticity (MPa) 60000 64100 60000 60000
Ultimate Elongation (%) 2.42 2.24 2.10 1.77
By comparing the two tables for V-Rod and V-Rod HM, the HM bar is easily
distinguishable by its increase in properties. Two factors contribute to the higher strength,
stiffness and elongation at rupture of HM bars. These include higher fibre volume and better
quality glass fibres. The V-Rod HM has much higher glass fibre contents at approximately
80% by weight while standard V-Rod has 77%. While E-glass fibres are used in V-Rods, the
fibre type in HM bars appear to be the higher grade S type as outlined in Table 1-2. The
manufacturer does not provide actual information about the fibre type in HM bars.
Trying to distinguish the two products visually can be difficult because of their
similar appearance. The two products are shown the photograph below (photo courtesy B.
Benmokrane)
14
Figure 2-2 Photograph of V-Rod and V-Rod HM (Courtesy B. Benmokrane)
The ComBAR is a reinforcing product manufactured by Schöck Bauteile GmbH of
Germany. The ComBAR bars are different from the other products because the surface of the
bar is milled, not sand coated. The ComBAR uses the typical E-Glass fibre and has a high
fibre weight ratio in excess of 87% by weight (Sheikh and Johnson, 2007). The properties of
the Schöck ComBAR are shown below in Table 2-8 (Weber 2007, Kiefer 2006).
Table 2-8 Mechanical Properties of Schöck ComBAR Reinforcing Bars
Nominal Diameter (mm) 8 12 16
Cross Sectional Area (mm2) 50.4 113 203
Tensile Strength (MPa) 1506 1365 1307
Modulus of Elasticity (MPa) 65900 67700 64000
Ultimate Elongation (%) 3.35 3.38 2.61
Schöck manufactures more sizes than just those listed in the above tables, the ones
shown above are just those that have been tested, audited and documented. The properties of
their largest bar (32mm) are discussed in section 6-1 in this thesis.
2.5.2 Glass Transition Temperature (Tg)
One property unique to FRP bars is the glass transition temperature. The glass
transition temperature is defined as the midpoint of a range of temperatures in which an
amorphous material changes from a brittle and vitreous state to a plastic state or vice-versa
(ISIS, 2006). Achieving the glass transition temperature does not mean the load carrying
15
capacity of the rod is exhausted; it is merely when the material begins changing form. More
information on the heat resistance characteristics of GFRP rods is provided in section 4.3.
Measuring the glass transition temperature involves a technique called differential
scanning calorimetry (DSC) where the sample is essentially heated and the energy required
to raise the temperature is recorded. When the glass transition temperature range is reached,
the energy requirement spikes and the corresponding temperatures can be recorded. Another
technique known as dynamic mechanical analysis (DMA) can also be used; however the
results from DMA testing differ slightly from DSC.
A higher glass transition temperature generally translates into a better product. It is
also possible to make the connection that a product with a higher Tg is more resistant to
chemical attack and is thus more durable. Glass transition temperatures are generally
undisclosed information. An example of the glass transition temperatures for the Schöck
ComBAR 12 and 16mm bars are shown in Table 2-9 below. All tests were done by Dynamic
Mechanical Analysis (Ehrenstein 2007, Schmachtenberg 2007).
Table 2-9 Glass Transition Temperatures for Schöck ComBAR Bar Size Number
of Tests
Lowest Point of
Temperature Range (oC)
Average Glass Transition
Temperature (oC)
Standard
Deviation
12 15 137 141 3.14
16 15 119 141 21.6
ISIS Min - - 110 -
2.5.3 Cure Ratio
Curing is defined as “the process of causing an irreversible change in the properties
of a thermosetting resin by chemical reaction” (ISIS, 2007). Many factors can affect the cure
ratio of a GFRP bar; three of the major ones are resin type, curing temperature and fibre
content.
16
The cure ratio requirements in the certification document are the strictest of all the
requirements for certification (refer to Table 2-4 for requirements). The cure ratio, similar to
the glass transition temperature, provides insight into the quality of the product which can be
used to infer some level of chemical resistance or durability without the direct testing of such
properties.
The requirements of the document specifically make no reference to the resin type for
the product being tested. It is well known that curing ratio for optimum properties of resin
vary with the types of resin (fib 2006). The guidelines should thus have separate
requirements for the three resin types outlined in section 2.2, although it is unlikely that any
manufacturer is using a pure polyester resin. Having different guidelines would better reflect
the different resin types and their curing behaviour. Some manufacturers have experienced
that for the more chemically resistant vinyl-ester type resins it is more difficult to achieve
high cure ratios than for the simpler polyester resins (fib 2006).
Testing done in the 1990s on the curing degree of a Bisphenol-A (BPA) based epoxy
resin of which the newer resins are derived from, indicates that some practical level of curing
exists of which attempting to cure beyond this limit is impractical (Guibe and Francilette
1996). The authors of the study go on to show that for a 50oC curing temperature the
limiting cure or conversion ratio is approximately 85-90% (Guibe, 1996).
The results of the 1996 study also correlate well with testing on current bars in which
experts in Europe have concluded that 95% is an excellent level of curing for vinyl-ester
resins and at those ratios, the bar is “more or less fully cured” (Ehrenstein 2007).
The issues with over-curing of resins become further compounded with the fact that
resins reach an optimal mechanical state at cure ratios less than 100%. Shown below in
Figure 2-4 is a plot of the resin and laminate strength against the cure ratio (Hülder 2008).
17
Figure 2-3 Tensile Strength vs. Cure Ratio for configurations of resin and laminates
(Hülder, 2008)
The key issue from the above chart is that there seems to be a peak cure ratio value
for achieving the optimal mechanical performance of the resin or laminate, exceeding that
value seems to cause a reduction in tensile strength. Some resins at higher cure ratios behave
in a more brittle manner without displaying any significantly improved properties (Hülder
2008). Thus, achieving the maximum possible cure ratio might be desirable, yet cure ratios in
the lower 90% range does not necessarily correlate to an inferior product.
The cure ratio for most of the bars available on the market range from 91% at the
lower end to 99%, which appears reasonable for the variety of resins and curing conditions
found in all the bars. The cure ratio requirements are currently under review in the draft CSA
certification document with the expected result being the relaxing of the current ISIS cure
ratio requirements.
The cure ratio also determines the durability designation of the GFRP products (ISIS
2007). While it is agreed that the cure ratio can provide insight into the chemical resistance
resulting in some idea of the overall durability of the FRP product, direct testing of the
alkali/chemical resistance is the most direct indication of the material performance. Such
tests are also required for certification and helps define the durability designation. In
addition, the durability designation is linked to the glass transition temperature which is
essentially making the same inference into durability from a material property as the cure
18
ratio does. The cure ratio should stay as a test requirement for certification and linked to the
type of resin, the durability designation’s dependence on the results of the cure ratio test
needs to be critically reviewed.
2.5.4 Other reinforcement products
In addition to the 4 different types of bars discussed above, some other FRP products
are emerging onto the market. Re-Bars DO Brazil have developed new FRP reinforcing bars,
one of them being a GFRP bar which is currently undergoing testing. Another emerging
product in Europe is made by a company known as STO Scandinavia which is currently
developing CFRP bars, it is unclear whether or not GFRP will follow soon after. Carbodur
CFRP rods manufactured by Sika Inc, are also being sold in Canada and it is also unclear
whether or not GFRP will follow soon after. None of these products are dealt with in any
detail in this study.
2.5.5 Summary
In this chapter, three different GFRP manufacturers and their products were discussed.
By comparing their mechanical properties it was observed that large differences exist
between products. Current GFRP products can essentially be classified into two categories,
high strength and normal strength depending on the ultimate tensile strength. These
differences stem from the different types of fibre and epoxy materials used which were
outlined in the chapter.
The cure ratio requirements for certification (ISIS 2006) are similar for all resin types
which cure in completely different manners and found to be too strict in some cases. This
fact points to a need for the reviewing of the durability designation’s dependence on the cure
ratio results.
19
3 LITERATURE REVIEW
3.1 Previous Work on the Flexural Behaviour
Some of the first published studies in FRP were military documents studying the use of
FRP in prestressed concrete (Wines and Hoff, 1966). Nawy and Neuwerth in the 1970’s then
tested 10 beams reinforced with GFRP bars in four point bending to investigate the flexural
behaviour. Further tests were conducted throughout the 1980s and early 1990s on the flexural
behaviour of GFRP reinforced concrete.
From the late 1990s onwards, much of the research is focused on trying to predict the
load deformation response of GFRP reinforced concrete. While there have been many tests
over the last 30 years, shown in Table 3-1 are 195 selected tests that used various concretes
and GFRP bars. Highlighted tests are covered in more details in the paragraphs to follow.
20
Table 3-1 Selected Flexural Behaviour Tests on GFRP RC
*** - Sections contained both steel and GFRP bars.
Reference Year Number of Specimens
Depth (mm)
Concrete Strengths
(MPa)
Reinforcement Strength (MPa)
Reinforcement Ratio (%)
Nawy, Neuwerth
and Phillips
1971 20 180 27.6 - 35.4 1069 0.19 - 0.45
Nawy and Neuwerth
1977 14 300 24.1 - 40.7 1069 0.696 - 2.54
Gangarao and Faza
1991 6 300 28.9 - 51.7 552 – 896 0.86 - 2.49
Brown and Bartholomew
1993 6 152.4 35.9 896 0.38
Brown and Bartholomew
1996 2 152.4 35.9 552 1.23
Al Salloum, Sayed and
Almusallam
1996 2 157 - 211
31.3 696 – 882 1.2 - 3.6
Benmokrane, Challal, and Masmoudi
1996 10 300 45 – 52 776 0.42 - 2.15
Sonobe et al. 1997 1 300 76 540 1.26 Al Salloum, Sayed and
Almusallam
1997 4 190.5 35.4 - 36.5 885 1.33
Zhao, Pilakoutas and
Waldron
1997 4 228.6 30 - 39.8 1000.7 1.27
Thierault and Benmokrane
1998 12 180 46 – 97 776 1.16-2.77
Vijay and Gangarao
1996 4 304.8 31 – 45 558 – 586 1.01 - 1.97
Pecce et al. 2000 3 185 30 770 0.96 Toutanji and
Saafi 2000 6 300 35 695 0.52 - 1.10
Yost and Goodspeed
2001 18 - 28 - -
Abdalla 2002 7 250 30 – 35 692 – 746 0.4 - 1.5 Yost, Gross and
Dinehart 2003 48 184 -
286 38-79 408 – 740 1.2 - 4.3
Nam, Lee et al. 2006 4 290 37 1205 *** Barris, Torres et
al. 2008 24 190 32 – 45 1353 0.99 - 2.66
21
3.1.1 Nawy et al 1971, 1977
The first major studies into the behaviour of GFRP reinforced concrete were done by
Nawy and his colleagues. A set of GFRP bars specifically manufactured for the research
program by American Cyanamid Co. were cast into concrete beams and tested in four point
bending.
As it was the first major study on the subject some major conclusions were drawn
including most importantly that the fundamental behaviour of GFRP reinforced concrete was
similar to that of traditional steel reinforced concrete. The authors concluded that their
current techniques for working stress design in steel were applicable to GFRP. An important
observation was made that many of the beams originally designed to fail in flexure failed in
shear.
While they noted that there was promise in using GFRP as a reinforcing product, they
also made the observation that because of the low stiffness of the bars, deflections and crack
widths need to be well controlled in GFRP reinforced concrete.
3.1.2 Brown and Bartholomew 1993
Brown and Bartholomew published their preliminary study into the flexural
behaviour of GFRP reinforced concrete in 1993. Their study also included the evaluation of
the bond behaviour of GFRP reinforcing bars. Regarding flexural behaviour, the authors re-
affirmed the conclusions drawn in the 1971 study by Nawy et al. One additional contribution
came from their bond investigation in which they noted that the bond mechanisms were
similar to steel but the bond strength was approximately 2/3 that of steel bars.
The final conclusions drawn from the study were that the flexural capacity and
behaviour can be well predicted but the both deflection and anchorage needs to be addressed
in design of GFRP reinforced concrete.
22
3.1.3 Benmokrane, Challal and Masmoudi 1996
This study, conducted at the University of Sherbrooke was done on a set of small
beams (200x300x3300mm) reinforced entirely with GFRP flexural reinforcement. The most
significant finding of that study was that at service loads; the number, width and penetration
of flexural cracks were all greater than a similar beam reinforced with steel reinforcing,
which went against conventional thinking that the number of cracks and their widths was
inversely related. Theoretically this can be explained by the lower stiffness of the
reinforcing which would increase the overall strain on the member at service loads. The other
conclusions of the study regarding the flexural strength and predictability of strength were
similar to those of the previous two studies outlined in this section.
3.1.4 Vijay and Gangarao 2001
After a lot of testing in the 1990’s was completed, Vijay and Gangarao compiled a
database of the tests and analysed them in order to help understand how the flexural
behaviour directly relates to the mechanical properties of the reinforcing bars. They
identified that the preferred failure mode of GFRP RC in flexure is concrete crushing
because the compression induced failure provides better deformability at ultimate conditions
as well as less deflection and crack widths at service load levels. While the benefit of better
deformability is true because of the nature of compression failure, reduced cracks and
deformation at service loads is primarily the effect of over-reinforcing with GFRP bars,
something that is necessary to ensure compression failure.
In another related study, Bakis et al. (2002) went on to further identify that
confinement of the compression zone in these over-reinforced members will increase the
ductility of the concrete and further enhance the overall ductility of the member. Thus based
on the conclusions from the Vijay and Gangarao, it was concluded by Bakis et al. that the
ductility of GFRP RC beams cannot rely on the inelastic behaviour of reinforcement as is the
traditional thinking with steel bars. The concrete as well as the amount and detailing of
reinforcement play vital roles in providing deformability. Vijay and Gangarao have identified
that the ductility of GFRP RC depends upon the following factors:
23
• Uniform elongation of FRP bars at different locations compared with localized steel
yielding.
• Uniform crack location and spacing in the case of FRP concrete beams.
• Bond between the bar and concrete.
• Plastic hinge formation in the concrete.
• Frictional force development along diagonal and wedge cracks.
Another major contribution was the further development of the concept of using
strain energies to determine the ductility or performance of a GFRP reinforced member. The
initial concept of using strain energies in FRP bars to determine the ductility was first
developed by Naaman and Jeong in 1995 and Jaeger, Tadros and Mufti in 1996. Those two
studies today even define the performance (J-Factor) of FRP-reinforced concrete in the
current 2006 Canadian Highways Bridge Design Code (CHBDC) for ductility. More
information on the J-Factor can be found in Chapter 7.
3.1.5 Yost, Gross and Dinehart 2003
In 2003 at the University of Villanova, a study was undertaken to determine how
accurately Branson’s equation (shown below) for effective moment of inertia predicts the
flexural stiffness of a GFRP reinforced member.
1 (3.1)
At the time, the current ACI 440 provisions for calculating the effective moment of
inertia used a correction factor based on the bond properties of the GFRP reinforcing bar.
Those provisions were developed based on the work by Gao and Benmokrane 1998. The
correction factor’s dependence on bond was later abandoned in the 2006 ACI 440 code in
exchange for dependence on the reinforcement ratio. Yost et al. noted that the transition from
gross to cracked section properties at flexural cracking were 8-10 times faster in GFRP-
reinforced beam than in a comparable steel reinforced beam. Depending on the
reinforcement arrangement, the transition borders on being instantaneous.
24
Their major conclusions were that the idea of a transition from gross to cracked
section properties is correct however the transition predicted by the Branson equation is too
slow. Using a cubic power in the equation for effective moment of inertia provided too slow
a transition. Work is still ongoing to determine a more suitable method as the modified
Branson is still the current ACI 440 code provision. Also, the paper included a performance
based design methodology called the EFS (Energy Factor of Safety) design method based on
limiting strain energies. More information on this method is covered in chapter 7 relating to
the performance design of GFRP reinforced members.
3.1.6 General conclusions on the flexural behaviour
None of the over 190 test specimens summarized in Table 3-1 is over 305 mm (12”)
in depth. Unlike the shear behaviour, bending behaviour is not known to be affected by the
member size and for that reason the general conclusions of the above studies related to
flexure are relevant to the large beams. However, it is very important to note that larger
sections are reinforced with larger bars and behave differently from their smaller
counterparts and their effects on the overall behaviour should be investigated. Even steel bars
have been known to have a size effect on their performance, and GFRP is no different. The
largest difference between the larger and smaller bars is the bond development requirements
as failing the larger bars require a significantly larger tensile force. This fact leads to the
bond development characteristics of the bars playing a paramount role in the structural
behaviour.
3.2 Previous Work on Bond
Bond development of reinforcing bars is becoming ever more important with the
rapidly increasing tensile strengths of the bars like ComBAR or V-Rod HM. The general
consensus from research is that the bond strengths developed in GFRP bars are lower than a
steel bar under identical conditions (Brown 1993). The reason for the lower bond strength
however is highly disputed. Outlined in this section is some important research conducted to
investigate the bond development of FRP reinforcing bars.
25
3.2.1 Malvar 1995
Malvar (1995) published one of the first studies with GFRP reinforcing bars in bond.
He tested various bars with different surface treatments in pull-out tests. Some of the major
conclusions of the study include that deformations on the surface of at least 5.4 percent of the
bar diameter can provide bond strengths up to 5 times greater than the concrete tensile
strength. That research indicates that the bond strength of the reinforcing is related to the
surface texture.
Malvar also identified the beneficial effect confinement has on the bond strength.
Confinement was determined to potentially increase the bond strength by a factor of three.
Also for the FRP bars tested Malvar determined that under identical conditions, similar steel
bars will have 1.2 to 1.5 times the bond strength of GFRP bars.
3.2.2 Tastani and Pantazopoulou 2002
In 2002, Tastani and Pantazopoulou evaluated the adequacy of the pull-out bond test
as well as further researching the effects of confinement and test setup on bond strength. The
major conclusion drawn from the study was that the beneficial effects of confinement are
very significant.
They went on to further demonstrate that experimentally determined bond strengths
are highly dependent on the test setup and type. Bars in direct cubic pull-out tests displayed
bond strengths 3 times greater than a similar bar in a pull-out specimen with less inherent
confinement. The reason for the increased strength was that in traditional pull-out tests, the
concrete is in pure compression. Other tests were conducted with concrete in tension and
provided more realistic estimates of bond strengths for reinforcement in beams, as in the case
of tension reinforcement in flexural members, the surrounding concrete is often cracked and
in tension. Tastani and Pantazopoulou used their results to demonstrate the large difference
in bond strengths from real world specimens and lab pullout specimens, a fact which they
indicate is highly dependent on the test setup and resulting beneficial effects of confinement
on the reinforcing bar.
26
3.2.3 Achillides and Pilakoutas 2004
Achillides and Pilakoutas tested over 130 cube pull-out specimens with various FRP
bars. By comparing smooth to ribbed bars, they made the conclusion that the bond strength
is heavily dependent on the surface deformations of the bar. They also note that one of the
primary modes of failure for FRP bars in pull-out bond is failure of the adhesive interface
between the resin matrix and glass fibres and not chemical adhesion between the bar and
surrounding concrete. That failure in their paper was noted as a shear lag failure which they
theorized as being one of the significant factors in determining the bond strength of bars for
concrete cube strengths greater than 30 MPa. Other researchers have also noted the shear lag
effect in their studies. The idea stems from their idealized normal stress distribution across
the bar cross section which they believe to be parabolic; their diagram is shown below:
Figure 3-1 Idealized normal stress distribution (Achillides, 2004)
The researchers also identified that in larger bars the difference between the
maximum and minimum stresses increases which causes premature failure of the bar in bond
caused by the orthotropic nature of the bar. These conclusions were based on post-failure
observations of the bar in which pulverized resin and fibres were noted along the failure
plane of the bar. The effect of size and interlaminar shear strength on the bond behaviour has
been contended by other researchers. Subsequent discussions in the journal (Wang 2004)
have indicated that steel bars (which are isotropic) also have a size effect on bond strength
but no shear lag. Other researchers believe that the issue stems mainly from the higher
possibility of defects in larger bars and not shear lag (Wang, 2004).
27
The researchers also concluded that in the case of chemical bond adhesion between
the bar and surrounding concrete, the bond strength seems to solely be dependent on the bar
diameter and not the concrete strength or surface deformations.
3.2.4 Wambeke and Shield 2006
Wambeke and Shield conducted a major study into the bond strength of FRP bars in
beams. These beams were representative of beams in real structures and not just convenient
idealized beam tests or eccentric pull-out tests. Through a large number of beam tests and the
compiling of a large database of previous work, Wambeke and Shield investigated the
current equations for the development length of FRP reinforcing bars in ACI 440 and many
other codes. The major finding of the study was an equation to predict the development
length required to avoid splitting or pull-out failure which in SI units was determined to be:
.
. . . (3.2)
Where db is the bar diameter, ffu is the tensile strength of the bar, fc’ is the concrete
strength, c is the clear cover to the bar, and α is a bar location factor. The researchers also
noted that the effects of transverse reinforcing (confinement level) are not accounted for in
this equation as more data is required. What is important to notice in the above equation is
that the surface treatment of the bar does not play any role in determining the development
length of the bar.
3.2.5 Mosley, Tureyan and Frosch 2008
Mosley at al. tested the bond strength of steel and various FRP reinforcing bars in
beam specimens. Their setup was a modification to a standard 4 point bending test; a total of
12 different beams were tested.
This study was parametric in nature involving bars of different stiffness and surface
treatments. The results of this research program could be used to directly compare the
different bars and identify the primary factors which affect the bond strength. Based on their
28
tests, Mosley et al. concluded that the primary factor that determines bond strength is the
axial stiffness (E) of the bar. They noted that the different surface treatments had an
insignificant effect on the bond strength which correlates well with Wambeke and Shield’s
proposed development length equation above.
Mosley et al. did not provide any mathematical relation between bond strength and
axial stiffness. Further research is needed to identify the analytical relationship between axial
stiffness and bond strength.
3.2.6 General conclusions on the bond behaviour
In general it was shown that the mechanisms of bond development are similar to that
of steel bars; however, the bond stresses are generally lower in the case of GFRP. It was
shown that the potential reasons included the lower modulus of elasticity of GFRP bars and
not necessarily the surface treatment. Some researchers have also theorized that a shear lag
effect exists in large bars as they develop large tensile forces in which premature
interlaminate shear failure occurs between the fibres in the bar. As is the case with steel bars,
research has shown that factors like confinement and concrete strength can significantly
affect the bond stresses developed between GFRP bars and concrete. In addition, bond
strengths determined via pull-out testing are not comparable with beam pull-out tests and
will not provide realistic estimates of real world bond strengths.
3.2.7 Summary
In this chapter, previous work on the behaviour of GFRP reinforced concrete was
discussed. In terms of the flexural behaviour, GFRP was shown to not fundamentally change
the behaviour of reinforced concrete. Some of the major conclusions regarding the flexural
behaviour are that the number, width and penetration of flexural cracks are all greater,
largely due to the low modulus of elasticity.
As well, because of the low modulus, issues such as deflection and crack widths need
to be well controlled. When evaluating the ductility of a GFRP section it was also determined
that because of the elastic to fail nature of the bar, the traditional definition of ductility does
29
not apply, hence several pseudo-ductility measures were developed which compare energy
states at ultimate flexural failure and some limiting condition based on service criteria.
Previous work on the bond development characteristics of GFRP have also shown
that in general the bond strength for a GFRP bar are lower than a comparable steel bar. The
primary reason for the lower bond strength is currently not agreed upon as there are studies
in support of either the surface treatment or modulus of elasticity as being dominant factors.
Research is ongoing in that field. Finally, because of a shear lag effect, previous bond
research has indicated that large GFRP bars will fail prematurely due to an interlaminar shear
lag effect.
30
4 DURABILITY OF GFRP REINFORCEMENT
Significant doubt still exists in the engineering community about how durable GFRP is
to environmental and long-term effects. Furthermore, in the case of glass FRPs, the material
is generally perceived as being significantly less durable than other materials like basalt or
carbon FRPs. The following sections deal with the durability of GFRP reinforcing rods in
several major areas, which are as follows:
• Resistance to Alkalis
• Effects of Creep and Sustained Load
• Response to extreme temperatures
• Fatigue behaviour
4.1 Alkali Resistance of GFRP Reinforcing Rods
While GFRP rods are corrosion resistant, they are not resistant to all forms of chemical
attack. The glass fibres that comprise the bar are made primarily from silicon and oxygen
(refer to chapter 2) which are highly susceptible to chemical attack from basic hydroxyl ions
(OH-). This section presents information on the alkali resistance of the commercially
available GFRP products.
4.1.1 Alkali resistance and testing
Alkali resistance in GFRP bars is achieved in two ways, one being the resin and the
other being a varnish applied on the outer surface of the bar. By coating each of the fibres,
the resin stops the penetration of the alkali ions. Varnishing the outside surface of the bar
with a different impregnable resin prevents ion ingress into the bar as a whole. Tests have
also shown that while the varnish prevents the ingress of ions, it increases the overall water
absorption of the GFRP bar as the varnish takes in more water than a bare bar (Ehrenstein
2007).
31
The resistance of GFRP rods to alkali is tested by submersion in an alkali solution for
an extended period of time. After exposure, the residual strength of the rods are determined
through testing and compared to a control sample or tensile tests on the same batch or lot.
The method prescribed by CSA (CSA S806-02, Annex O) which was modified by ISIS
(2006) requires submersion in solution for 3 months at 60oC. The chemical composition of
the alkali solutions is:
118.5g [Ca(OH)2] + 0.9g [NaOH] + 4.2g [KOH] + 1L [H2O].
European testing standards require a special Masthoff solution where the
concentration is 10 times greater to represent a harsher environment more representative of
the actual pH and concentration of concrete pore water solution. The chemical composition
of Masthoff Solution is:
1185 g [Ca(OH)2] + 9g [NaOH] + 42g [KOH] + 1L [H2O].
After submersion, the bar is tested in direct tension in a standard tensile test.
Sustained stresses can also be applied during submersion depending on the test and the
intended use of the bars. Some researchers also choose to test the alkali resistance of the bar
while embedded in wet concrete prisms. A report published by fib Task Force 9.3 (fib, 2006)
states that significant research has noted that testing with aqueous alkali solutions is far more
aggressive than testing in wet concrete prisms because of the increased mobility of the OH-
ions in the aqueous solution. Thus results from accelerated aging tests can vary depending on
the alkali media and test method used.
4.1.2 Alkali resistance of commercially available bars
The results from selected tests on three types of bars either from research programs
or reported by the manufacturers are summarized in Table 4-1. All of the listed tests were
carried out in simulated solutions, not concrete prisms (Dejke 2002, Nkurunziza 2005,
Weber 2005).
32
Table 4-1 Summarized Alkali Resistance of Commercially Available GFRP bars. Manufacturer &
Bar Bar Size
(mm)
# of Tests
Sustained Stress (MPa)
Exposure Time (hrs)
Exposure Temp. (oC)
Residual Strength (MPa)
Strength Retention
(%) Pultrall V-Rod
9.5 5 157 10,000 RT 555 84
Pultrall V-Rod
9.5 4 239 10,000 RT 429 65
Schöck ComBAR
16 8 0 2,000 + 60oC 1101 84
Schöck ComBAR
16 6 250 2,000 + 60oC 1024 79
Schöck ComBAR
16 7 300 2,000 + 60oC 1103 84
Schöck ComBAR
16 7 350 2,000 + 60oC 1149 88
Hughes Brothers ASLAN
NS NS 0 2,400 +60oC NS 65
Hughes Brothers ASLAN
9.5 5 0 2,160 +57oC 417 60
Hughes Brothers ASLAN
9.5 7 0 1680 +60oC 493 64
(NS – Not Specified in Report, RT – Room Temperature)
The test results listed above in Table 4-1 are only a snapshot of the research into
alkali degradation. A significant amount of additional research has been done, however much
of that work was done on bars that are either not currently available or on older generation
products of the major manufacturers.
Based on the results presented in Table 4-1, making any distinctive conclusions about
the alkali resistance of GFRP bars in general is difficult as not only do the products differ,
the test methods used do as well. Currently, no internationally accepted unique standard
exists for the testing of the alkali resistance. Furthermore, because the fibres and resins used
in the products are trade secrets, making any conclusions based on resin or fibre material is
difficult.
33
4.2 Creep Rupture Strengths
4.2.1 Creep rupture test method (CSA S806‐02)
Creep rupture tests are used to determine how well the material behaves under
sustained load conditions. Bars are to be subjected to a sustained tensile load for a set time.
The load level is held on the sample until failure. After failure, the load and corresponding
time to failure are reported. Loads and failure times that cross three epochs of time (10, 100,
1000 hours) are recommended. The points are then plotted and the extrapolated millionth
hour creep strength is determined from regression analysis which is often referred to as the
endurance limit of the FRP material. Minimum R2 curve fit parameters are prescribed for
acceptable predictions.
While the method prescribed in the Canadian code (CSA S806-02) is direct tension
without the presence of alkali, some testers choose to incorporate alkali. One such test with
alkali and sustained load was a test done on the ComBAR bars. This test used bars cast in
wet concrete (Figure 4-1). The bars were cast into a concrete prism and two concrete blocks.
Hydraulic jacks were then placed between the blocks to provide the tensile load. The time to
failure and the failure load were recorded.
Figure 4-1 Setup for Creep Rupture Tests on ComBAR bars (Weber, 2005)
34
4.2.2 Creep rupture strength of Available GFRP bars
Weber (2005) completed the above mentioned tests on three of their bar sizes, 8, 16
and 25mm. Three different temperatures were maintained during the experiments as well. It
should be noted that the Canadian code (CSA S806-02) only specifies an ambient
temperature test. Listed below are the results from the creep rupture tests for the 16mm
ComBAR bars for all the different temperatures. The time to failure for various sustained
stresses, shown in the table, are plotted on a logarithmic time scale in Figure 4-2.
Table 4-2 Summarized Creep Rupture Data for 16mm ComBAR bars
Ambient Temperature (30oC) Elevated Temperature (40oC) High Temperature (60oC) Failure Stress
(MPa) Time (Hours) Failure Stress
(MPa) Time (Hours) Failure Stress
(MPa) Time (Hours)
1160 164.7 1000 378 1074 46 1160 119.7 960 574 1044 39 1140 309.25 920 1414 1000 125 1125 313.9 900 2355 1000 192 1125 885.5 880 2762 975 130 1100 1753 850 2853 975 132 1100 319 - - 930 132 1000 2709 - - 930 93.5 1000 3275 - - 880 139 950 5021 - - 880 195 940 3288 - - 800 582 900 6559 - - 800 547 900 2876 - - 750 1099
- - - - 750 1625 - - - - 700 3067 - - - - 683 2445 - - - - 683 2760 - - - - 650 6700 - - - - 650 3988
A regression analysis for each of the temperature ranges yielded the relationships
between time to failure and the sustained stress that are shown in Table 4-3. The R2 value
for each equation is also shown in the table. By evaluating these relationships for a time of 1
million hours, the millionth hour creep strengths can be predicted. The 3 creep strengths
along with the minimum requirement are listed in Table 4-4 below.
35
The millionth hour creep strength in this paper differs slightly from the value
published in the compliance report for the 16mm bar (Sheikh and Johnson 2007) because of
the inclusion of a few more test points and a better regression analysis result.
Figure 4-2 Plot of Sustained Stress vs Failure Time in Creep Rupture Tests
Table 4-3 Linear Regression of Creep Rupture Data for 16 mm Bar
Temperature Equation Limitations R2 Ambient (30) Sustained Stress = -64.9 Ln(Time) +
1499.5 Time must be in hours and greater than 200.
0.935
Elevated (40) Sustained Stress = -61.12 Ln(Time) + 1358.2
Time must be in hours and greater than 400.
0.937
High (60) Sustained Stress = -84.84 Ln(Time) + 1363.1
Time must be in hours and greater than 100.
0.809
30oC
40oC60oC
36
Table 4-4 Predicted Millionth Hour Creep Strengths for 16mm Bar
Temperature Millionth Hour
Creep Strength (MPa)
Millionth Hour Creep
Strength as % of UTS
Ambient (30oC) 602 46.1
Elevated (40oC) 513 39.2
High (60oC) 191 14.6
ISIS (2007) Requirement
(Ambient 30oC) 457 35
All the tests on the 8mm and 25 mm bars were conducted at 60oC. The results are
shown in table 4-5 below.
Table 4-5 Creep Rupture Data for 8mm and 25mm ComBAR bars
8mm ComBAR (60oC) 25mm ComBAR (60oC)
Failure Stress (MPa) Time to Failure
(Hours)
Failure Stress (MPa) Time to Failure
(Hours)
980 72 1000 54
950 119 962.7 234
925 190 950 293
900 256 925 290
850 526 900 425
800 945 850 811
760 1540 800 812
740 2047 750 980
700 3526 725 2347
700 945 700 3768
- - 650 6185
37
The high temperatures used in the tests render making a millionth hour prediction for
ISIS requirements erroneous because of how extreme the exposure was. The results in Table
4-4 show how significant an effect the highly elevated temperatures have on the bar
behaviour. By comparing the data from three sizes at 60oC, it is possible to determine if there
is any significant size effect between the three sizes. The results from the creep rupture tests
conducted at 60oC on all three sizes (8, 16, 25mm) are plotted in Figure 4-3. The data from
all three sizes follow a very distinct linear pattern and have very good agreement with one
another. It can thus be concluded that bars of different sizes behaved very similarly.
Figure 4-3 Comparison of ComBAR sizes in creep rupture
4.3 Performance in Extreme Temperature Environments
4.3.1 Glass Transition Temperature (Tg)
As outlined in section 2.5.2, the glass transition temperature (Tg) is defined as the
midpoint of a range of temperatures in which the glass fibre changes from a brittle to a
vitreous state (ISIS, 2006). There are two acceptable methods for measuring the transition
temperature: Dynamic Mechanical Analysis (DMA) and Differential Scanning Calorimetry
Creep Rupture Data for Schock ComBAR(Exposure Temperature 60oC))
0
200
400
600
800
1000
1200
10 100 1000 10000
Time to Failure (Hours)
Sustained Stress (M
Pa)
8mm
25mm
16mm
38
(DSC). The glass transition temperature is primarily determined by a combination of the
materials used in the bar and the degree of curing of the epoxy resin.
Shown in table 4-6 again are the measured glass transition temperatures for two sizes
of ComBAR bars from section 2.5.2. (Ehrenstein 2007, Schmachtenberg 2008). Data from
the other two manufacturers were not available or provided. The minimum glass transition
temperature for DMA testing is 110oC for the high durability designation according to the
ISIS guidelines (2006). Other research has indicated that the glass transition temperature for
Pultrall V-Rod is around 120oC; however exact numbers were not provided (Robert et al.
2009).
Table 4-6 Summary of Glass Transition Temperature Results
Bar Diameter
(mm)
# of
Tests
Average Transition
Temperature (oC)
Lowest Transition
Temperature (oC)
Standard
Deviation (oC)
12 15 141 137 3.14
16 15 141 119 21.6
4.3.2 Bar mechanical property change under extreme heat
Because of the existence of the glass transition temperature, high temperatures can
have a significant effect on the mechanical properties of the GFRP rebar. High temperature-
induced reductions in mechanical properties will occur more rapidly once the glass transition
temperature is exceeded. It important to note that GFRP bars can maintain tensile strength at
temperatures well beyond the glass transition temperature as the resin material is the first to
deteriorate under high temperatures.
Tests conducted at the Brunswick Institute for Construction Materials Testing have
investigated the behaviour of the ComBAR materials under extreme heat. Shown below in
Figure 4-4 is a plot of the tensile strength change with temperature after the glass transition
temperature is exceeded.
39
Figure 4-4 Tensile Strength vs. Temperature for ComBAR bars (Nause 2005)
The significant reduction in tensile strength that begins at extremely high
temperatures (450oC in the above chart) occurs due of complete degradation of the resin
matrix. While the degradation on bar strength takes place at the temperature in the range of
400oC, that is not the case for GFRP reinforced concrete in general as internal bars are
typically heavily insulated by the surrounding concrete.
Research at the University of Sherbrooke has also investigated the effects elevated
temperature has on the properties of GFRP reinforcing bars. Pultrall V-Rod samples were
tested at various temperatures to determine their mechanical properties. The results of the
experiments are shown in Figure 4-5 below (Robert et al. 2009). A comparison of Figures 4-
4 and 4-5 shows similar adverse effect of high temperature on strength. However, significant
increase (about 20%) in strength at lower temperatures in Figure 4-5 is somewhat unusual.
40
Figure 4-5 Tensile Strength vs. Temperature for V-Rod bars (Robert et al. 2009)
4.3.3 Bond strength degradation under extreme heat
Bond strength degradation occurs instantaneously as the temperature rises resulting in
severe bond strength degradation by the time the glass transition temperature is reached. A
plot of the bond strength from both pull-out and push-through tests with temperature is
shown in Figure 4-6. The circular points represent a push through test while the triangular
marks a pull-out test. Failure at temperatures below Tg is due to splitting on the concrete,
while failures beyond Tg were due to decomposing of the resin and shearing off of the ribs of
the bar. As shown in the figure, significant drops in the peak bond strength can be noted for
both the pull-out and push-through tests at temperatures below Tg.
Severe deterioration of the bond strength with temperature even at lower range of
temperatures is a common trend for all GFRP reinforcing bars. Katz, Berman and Bank
(1999) also tested the bond strength of ASLAN FRP bars and found similar results.
41
Figure 4-6 Pull-out and Push-through bond testing at various temperatures (Weber 2008)
Shown below in Figure 4-7 are load slip charts for a helically wrapped sand coated
GFRP bar (Labelled CPH) and a deformed steel reinforcing bar (Labelled ST).
Figure 4-7 Load Slip Charts for ASLAN 100 FRP under varying temperature conditions.
(Katz et al. 1999)
42
Again it can be noted that temperature plays a significant role in determining the
bond strength. While the plot about the steel bars shows that steel is also affected by
temperature, the effect is far less profound than in the case of GFRP.
4.3.4 Response of GFRP Reinforcing Bars to Extreme Cold
While research on the effects of extreme heat is well documented, research work on
the effects extreme cold temperatures have on GFRP bars is limited. Fire conditions that are
similar to those simulated in experiments like those in section 3.3.4 and 3.3.5 are found only
in extreme events. In climates similar to Canada’s, cold temperatures as low as -40oC can be
found daily for more than one quarter of the year; it is for that reason that ISIS specifications
require evaluation of GFRP properties at low temperatures. Extreme low temperatures have
been theorized to cause matrix hardening, micro-cracking and the degradation of bond
between the fibres and matrix (fib, 2006).
Part of the experimental work in this research project was to conduct low temperature
tests on GFRP bars. COMBAR bars from three different sizes were tested at temperatures as
low as -40oC, the results of which and relevant analysis are presented in the next few
chapters. Pultrall Inc. have also had tests conducted on their product (Robert et al. 2009),
refer to Figure 4-5 for the results.
4.4 Fatigue Strength of GFRP Reinforcing Rods
4.4.1 Test method and results of fatigue testing
The determination of the fatigue strength of GFRP rods for the higher strength
products is a very difficult test to perform because of the nature of testing GFRP bars in
tension. According to CSA S806-02 standards, fatigue testing is done in direct tension
through cyclic tension. Because of the high strength of the bars and the need to grip the bars
with metal couplers, cycling in direct tension usually results in failure of the bars at the
coupler locations.
43
Tests done at the University of Karlsruhe have looked at the fatigue strength of the
ComBAR bars in cyclic four point bending tests. 16mm ComBAR bars were cast into T-
beams and cycled in flexure until failure. The section properties and test setup is shown in
Figures 4-8 and 4-9.
Figure 4-8 Specimens and Reinforcing for Dynamic Tests at Karlsruhe (Kreuser 2007)
For these tests, peak and minimum tensile bar stresses were selected and the beams
were cycled such as to produce the desired bar stress values. The number of cycles to failure
was recorded. All of the results are outlined in Table 4-7.
Figure 4-9 Test Setups for fatigue strength testing at Karlsruhe (Kreuser 2007)
44
Table 4-7 Cyclic Bending Tests Results (Kreuser 2007)
Test Number Peak Stress
(MPa)
Minimum
Stress (MPa)
Amplitude
(MPa)
Number of
Cycles Failure of bar
16/1 250 150 100 2,539,288 Yes
16/2 250 150 100 3,149,240 Yes
16/3 250 150 100 3,002,000 No
16/4 250 140 110 2,022,800 No
16/5 250 70 180 321,239 Yes
16/6 250 90 160 2,055,336 No
16/7 250 25 225 1,920,500 Yes
16/8 250 70 180 2,018,000 No
16/9 300 240 60 2,000,000 No
16/10 300 240 60 2,000,000 No
16/11 300 240 60 5,000,000 No
16/12 150 15 135 3,047,612 No
16/13 175 17.5 157.5 2,601,966 Yes
16/14 200 20 180 3,000,000 No
16/15 175 17.5 157.5 1,969,962 Yes
The tests have presented some interesting results, the first being that the number of
cycles to failure appear to be sensitive to the amplitude of the cycles more than the peak
stress itself. This is clear from looking at tests 16/9 to 16/11 which have the highest peak
stress yet no failure It should be noted that the bars display linear response until failure and
the maximum stress in these tests, 300 MPa, is less than 25% of the ultimate tensile capacity
of the bars. The results of Table 4-7 also show that GFRP bars are capable of withstanding a
high number of load cycles at relatively high stresses.
45
4.5 Do Simulated Lab Tests Reflect the True Conditions?
Determination of the long-term durability properties of engineering materials would
not be possible without accelerated testing. Accelerated aging tests like the creep rupture
tests and alkali resistance tests described in this chapter have shown that in as little as 2,000
hours, in the worst case scenario, significant degradation of the bars has occurred. While
these simulated and accelerated lab tests can give insight into the millionth hour behaviour in
less than 10,000 hours, the question remains as to just how accurate are they?
In 2007 members of the ISIS Canada Research Network conducted an in depth field
study of five GFRP reinforced concrete bridges between the ages of 5 and 8 years old (Mufti
et al. 2007). After using both destructive and non-destructive test methods they found that no
significant degradation of the GFRP reinforcing bars had occurred. Tests like Scanning
Electron Microscopy, Fourier Transform Infrared Spectroscopy and Energy Dispersive X-
Ray analyses showed no significant change or degradation between the control samples and
the samples taken from the real structures.
The results of the field study are encouraging however; in the overall lifespan of a
structure 8 years is not a very long time and it remains to be seen if the simulated lab tests do
indeed reflect the real world conditions for a structure 40 to 50 years old.
4.6 Summary of Durability
In this chapter the durability of various GFRP products was discussed. Topics
ranging from alkali attack and creep rupture to extreme heat and cold exposure were
discussed.
It was found that GFRP bars are susceptible to attack in strong alkali solutions. With
no internationally accepted standard, testing is currently done in a variety of different
manners with differing results. The alkali resistance of GFRP bars is dependent on the
following factors:
46
• Fibre type and quality of fibre matrix bond
• Resin type
• Varnishing of the outside surface
• Temperature of the test
• Mobility of the OH- ions in the accelerated test environment
• Level of sustained load
From the creep rupture tests, the endurance limit for one type of GFRP bars was
predicted to be approximately 46% of the ultimate tensile strength for a room temperature
exposure condition, which is above the ISIS minimum requirement for FRP bars. It was also
shown that the endurance limit decreases significantly as the sustained exposure temperature
increases. Based on a comparison of three different bar sizes under identical exposure
conditions it was concluded that there is no discernable size effect for creep rupture.
GFRP bars were shown to be greatly affected by high temperatures. This fact is
particularly true when the glass transition temperature is exceeded. It was also shown that the
degradation of bond strength between GFRP bars and concrete is particularly susceptible to
high temperatures and occurs well before the degradation of material properties.
Testing the fatigue strength of GFRP bars in direct tension is difficult because of the
propensity for failure to occur at the coupler level. From tests done in cyclic flexure the
fatigue capacity of the bar seems to be more dependent on the amplitude of the load than the
peak repeated stress.
Finally, while all of the lab tests give an indication of the durability of the bars,
evidence suggests that simulated lab experiments over-estimate the degradation that occurs.
Both non-destructive and destructive field tests of 10 year old in-service structures reinforced
with FRP bars show little to no degradation of the bars (Mufti et al. 2007).
47
5 EXPERIMENTAL WORK
The experimental work in this research program consists of two parts. The first is the
evaluation of the GFRP bar properties, while the second part relates to the testing of GFRP
reinforced concrete elements. The first phase of the experimental work involved testing small
diameter bars of sizes between 8 and 16mm in direct tension both at room temperature and
under extreme cold temperatures. The second phase involved testing the largest diameter bar
available in beams. Five samples with one single 32mm bar was embedded in a single
concrete beam, the beams were tested until failure. Details of both the test programs are
given in this chapter and results are presented in Chapter 6.
5.1 GFRP Extreme Cold Temperature Tests
5.1.1 Objective of cold temperature tests
Canadian winter represents one of the harshest exposure environments for any
concrete structure. When considering the effects a cold winter has on structures, it is
generally believed that the only significant problem is the amount of de-icing salts and their
effects on reinforced concrete. Cold temperatures in general have been shown to affect the
properties of materials in a significant manner and most importantly modulus of elasticity
and the ductility. In many parts of Canada, temperatures on some extreme winter days can
reach as low as -40oC on the windward side of bridges and buildings. It is for this reason that
the ISIS certification guidelines (2006) require testing at cold temperatures.
Because no previous work existed on the effect extreme cold temperatures have on
GFRP bars, specialized testing equipment and specially made test specimens had to be used
in the experiments. The following sections outline the details of cold temperature testing.
More detail regarding these tests can be found elsewhere (Sheikh and Johnson 2008).
5.1.2 Specimen preparation
Fifty specimens of ComBAR bars in three different sizes (8, 12 and 16mm) were
prepared for testing at -40oC. The number of samples and their dimensions are shown in
48
Table 5-1. Each specimens was given a designation TCBx-a where “x” denotes the bar
diameter in mm and “a” is the test number.
Table 5-1 Samples for Cold Temperature Testing
Nominal Diameter 8 12 16
Number of Samples 18 18 17
Cross Sectional Area (mm2) 50.3 101 201
Because of the high strength of the GFRP bars, testing in direct tension similar to a
standard steel reinforcing rod is not possible. Clamping directly onto the GFRP bar at the
ends would crush the fibres at high loads rendering the tests incomplete. It is for this reason
that bars were attached to steel couplers at ends with the help of epoxy as shown in Figure 5-
1.
Figure 5-1 Schöck ComBAR specimens with attached couplers
The bars were then preconditioned as per ASTM D618 – Conditioning of Plastics, a
method recommended in the ISIS Canada guidelines for testing GFRP bars at cold
temperatures. The bars were conditioned in the Thermotron Unit in the Mechanical Testing
Laboratory at Kinectrics Inc. The Thermotron unit uses Liquid coolant injection and cold air
to maintain temperatures as low as -35oC for extended periods of time. The bars were kept in
the chamber for anywhere between 24 and 96 hours with a few bars exceeding 96 hours of
conditioning. Figure 5-2 shows the Thermotron Unit.
49
Figure 5-2 Thermotron Unit used for preconditioning
When the bars were ready for testing they were moved individually in an insulated
transport unit made of extruded polystyrene. This method ensured that only a minimal
increase in temperature occurred during transport from the preconditioning location to the
test location.
5.1.3 Control sample testing at room temperature
Prior to testing any samples in the cold environment it was necessary to test control
specimens at room temperature against which the cold temperature test data could be
compared. In addition to the available tests already conducted (refer to chapter 2) additional
samples were tested in this study to check the variations from one batch to other. For the
room temperature tests; bar samples were mounted into the universal testing machine without
any preconditioning or special equipment. A direct tensile test was conducted on the samples.
A picture of the setup is shown below in Figure 5-3.
50
Figure 5-3 Overview of Control Sample test setup (16mm sample shown (TCB16-01))
A linear variable displacement transducer (LVDT) was installed to verify the
elongation of the sample. Because of the potential for a large amount of slip between the
jaws of the machine and the couplers to occur at the high loads required for failure, the
LVDT was also used to calibrate the jack position readings. Subsequent tests on smaller
diameter bars (8 and 12mm) used strain gauges instead of LVDTs as they proved to be more
accurate in determining the elongation of the bars.
The universal test machine shown in Figure 5-3 was the one used throughout this set
of experiments. This Satec/Instron machine has a capacity of 120,000 lbs (540 kN) and can
be operated with load or displacement control. The displacement rate can be varied between
1 mm and 10 mm per minute.
51
5.1.4 Cold temperature test setup
To follow the guidelines for testing at cold temperatures, it is necessary to maintain
the test environment at -40oC. This environment was achieved by using an environment
chamber which was specially designed to mount onto the SATEC testing machine. The
Bemco FTU5.5 environmental chamber uses liquid nitrogen and is rated to cool down to -
120oF (-84oC). Liquid nitrogen was provided with an XL45 LN2 tank. Figure 5-4 shows the
environmental chamber in the testing machine.
Figure 5-4 Universal Test Machine with Attached Environmental Chamber
Because the built-in thermostat on the Bemco Chamber was inadequate for use as it
was not calibrated before testing, a thermocouple was placed inside the chamber and the
temperature was monitored at all times during testing. The manufacturer of the thermocouple
unit was Barnant and the unit was calibrated prior to testing. Figure 5-5 below shows the
BEMCO control unit and the Barnant thermocouple readout on the top highlighted in red.
52
Figure 5-5 BEMCO Control Unit and Thermocouple Readout
5.1.5 Specimen mounting
While that the couplers at the ends of the bars made anchorage of the specimens in
the machine easy to achieve in general, using the thermal chamber provided new challenges.
The clamping devices of the universal testing machine could not function inside of the
testing chamber during testing; it is for this reason that the bar samples were mounted such
that the metal couplers were outside of the chamber walls exposed to the room temperature
air. This arrangement was beneficial to the testing as the thermosetting resin which connects
the couplers to the GFRP bar was left at room temperature. To ensure minimal losses of heat
through openings in the chamber for the ComBAR samples, the openings were insulated
using fibreglass insulation. Figure 5-6 shows the bar sample mounted inside the chamber for
testing with the fiberglass batt insulation.
It was imperative that the couplers be insulated against the cold temperatures of the
chamber. Since the coefficient of thermal expansion for the thermosetting resin used to bond
the couplers is greater than that the steel couplers, at low temperatures, the resin begins to
shrink faster than the couplers. It was believed that this shrinkage would result in lower bond
strength between couplers and ComBARs.
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54
5.1.7 Testing procedure
Testing was carried out at a set strain rate of 2mm per minute. The samples were
tested until failure in a monotonically applied tension load. Failure of the specimens either
occurred as a result of failure of the epoxy between the coupler and the bar or as a complete
bursting failure of the ComBAR itself. A graph showing axial load versus jack head position
were plotted simultaneously as the test was being conducted. This plot was made using the
calibrated SATEC testing machine.
5.2 GFRP-Reinforced Large Beams
5.2.1 Objective of the test As it was shown in Section 3.1, of all the flexural tests reported in the literature, none
of the specimens was deeper than 300mm. Flexure behaviour is not as significantly affected
by size as shear or bond behaviour are, however, the larger specimens are typically
reinforced with larger bars. The size effect of larger bars is documented and does exist.
In the specific case of GFRP RC, many current researchers dismiss large GFRP bars
as being impractical and unsafe for design. Reasons such as shear lag are commonly cited as
reasons for restricting the use of large bars. It is important that the behaviour of these large
bars in larger sections is well understood as manufacturers have been making bars with
diameters equal to or greater than 25mm. These large beam tests were developed primarily to
look at the overall tensile behaviour of the large reinforcing bars. The cracking, load-
deflection and bond behaviour of the entire reinforced concrete beam were also studied as
part of the test.
5.2.2 Specimen design
A total of 5 identical beam specimens were designed and constructed using 32mm
ComBAR bars as primary load carrying tensile reinforcement. The design of the beams is
shown in Figure 5-7.
55
Figure 5-7 Geometry of Large Beam Samples
The 5 beams were all rectangular in geometry with a width of 400mm and a height of
650mm. The corners of the beams were rounded at a 25mm radius such that they can be
wrapped with surface mounted Glass FRP. The only major variable in the experiments were
the shear span and the loading type.
5.2.3 Construction and casting of the beams The five beams were cast together in the formwork shown in Figure 5-8 . As it can
be seen in the picture and the design drawings, small steel bars were used in every beam to
create the reinforcing cage to house the GFRP bar. The smaller bars were U.S. #3 (3/8 in)
[9.53mm]. The steel bars were used as transverse reinforcement in the cages in order to not
only provide shear strength but also confinement of the compression concrete. The
mechanical properties determined by coupon tests for the #3 bars are shown in Table 5-2.
Table 5-2 Mechanical Properties of #3 Steel Reinforcing Bars Bar Size Area
(mm2) Yield Stress, (MPa)
Yield Strain,
Elastic Modulus,
(MPa)
Strain Hardening
Strain
Ultimate Stress, (MPa)
Ultimate Strain,
US #3 71.3 496 0.0025 198580 0.0283 605 0.1711
56
Figure 5-8 Formwork for the large beams
The specified 28 day strength of the concrete was 85 MPa. The beams were covered
with moistened burlap for the first 7 days. Removal of the specimens from the formwork
occurred 10 days after casting. FRP wrapping of the samples occurred at a much later date.
The concrete cylinder strength was found to be 67 MPa at time of testing of beams.
To ensure that adequate shear strength was provided, the beams were also wrapped in
two layers of TYFO S GFRP wrap provided by Fyfe Co. The additional layers of wrapping
also provide excellent confinement of the compression concrete which was also desired. The
mechanical properties of the TYFO S wrap were determined through coupon testing and
shown in Table 5-3 below. Figure 5-9 shows two of the beams that have been wrapped after
removal from the formwork.
Table 5-3 Mechanical Properties of GFRP Wrap
FRP Nominal
Thickness (mm)
Tensile Strength (MPa)
Rupture Strain
(mm/mm)
Elastic Modulus (MPa)
TYFO S GFRP 1.25 518 0.02031 25488
57
Figure 5-9 GFRP wrapped beams
5.2.4 Instrumentation
A combination of strain gauges and LVDT’s were used in this study to instrument the
specimens. To determine the strain in the GFRP bars during testing, three strain gauges were
mounted onto the GFRP bars close to midspan. The gauges were manufactured by Tokyo
Sokki Kenkyoju Co. (TML) and both 60mm and 5mm long gauges were used in this study.
The standard Cyanoacrylate glue was not used to bond the gauges to the GFRP bars.
Cyanoacrylate (commonly referred to as CN glue) needs water to set, typically the surface
moisture on the steel bar is sufficient for that reaction, however with GFRP there is no
surface water film on the bar after the varnish is removed. Instead, M-Bond AE10 glue from
MicroMeasurements Inc. was used with good results. None of the steel bars in the cage were
instrumented.
During testing, 5 vertical LVDT installed at the supports and the quarter points were
used to measure the vertical deflection of the beams. In the middle 600mm of the beams 3
horizontal LVDT’s were mounted at various depths of the beams to capture the horizontal
strains in the region of maximum moment. The results of those three LVDTs were used to
calculate the longitudinal strain gradient (curvature) at midspan.
58
5.2.5 Test setup and procedure
Testing was done at the Structures laboratories at the University of Toronto. The
specimens were tested in a large MTS testing machine. Shown in Figure 5-10 is the design of
the test setup. Figure 5-11 shows the test machine used in the experiments.
Figure 5-10 Test setup for large beam tests
Figure 5-11 MTS Machine used for testing
59
The position of the load points varied by the specimen in an attempt to determine the
effect the bonded length had on the GFRP bar behaviour. Three different setups were used in
the testing and they are outlined in Figure 5-12 below.
Figure 5-12 Load arrangements for large beam tests
Load was measured by the MTS testing machine and midspan displacement was
calculated using the LVDT measurements. Loading was applied monotonically at a rate of
approximately 0.25 kN/second until failure. All LVDTs and strain gauges were connected to
a data acquisition system with a sampling rate of 2Hz. Shown below in Figures 5-13 and 5-
14 are pictures of one of the beams before and during testing (TCB 3202). The results and
accompanying analysis are presented in the next two chapters.
60
Figure 5-13 Picture of TCB 3202 before application of load.
Figure 5-14 Picture of TCB 3202 during testing.
61
6 RESULTS AND DISCUSSION
The results of both experimental programs outlined in Chapter 5 are analyzed and
discussed in detail in this chapter. The cold temperature tests are discussed first followed by
beam tests.
6.1 Cold Temperature Test Results
A total of 55 test results are discussed in the sections to follow. The general
observations from the tests including types of failure are presented first. Analysis and
discussion of the results follow behind.
6.1.1 Results of the tensile tests on 8mm bars
The test results of all 18 tests conducted are shown in Table 6-1. It is important to
note that the modulus of elasticity was calculated directly from measurements from adhered
strain gauges. Tests with no modulus or elongation data indicate that the strain gauge broke
or malfunctioned during testing. A total of 14 of the 18 (78%) samples ruptured while the
remaining ones had a de-bonding failure between the coupler and the bar. In Table 6-2 the
results of the 14 tests that ruptured are summarized.
62
Table 6-1 Results of 8mm Cold Temperature Tests
Ultimate Strength (MPa)
Modulus of Elasiticity
(MPa)
Ultimate Elongation
(%)
Failure Type Notes/ Remarks
TCB 8-01 1379 59072 2.34 Rupture TCB 8-02 1291 63022 2.05 Rupture TCB 8-03 1388 53458 2.60 Rupture TCB 8-04 1132 61253 1.85 Coupler Slip TCB 8-05 1415 66706 2.12 Rupture TCB 8-06 1362 70934 1.92 Rupture TCB 8-07 1194 56570 2.11 Coupler Slip TCB 8-08 1423 67651 2.10 Rupture TCB 8-09 1149 64681 1.78 Coupler Slip TCB 8-10 1370 73560 1.86 Rupture TCB 8-11 1344 60520 2.22 Rupture TCB 8-12 1088 - - Coupler Slip TCB 8-13 1379 66706 2.07 Rupture TCB 8-14 1353 59482 2.27 Rupture TCB 8-15 1371 65886 2.08 Rupture TCB 8-16 1371 61133 2.24 Rupture Control (RT) TCB 8-17 1406 56570 2.49 Rupture Control (RT) TCB 8-18 1346 62265 2.16 Rupture Control (RT)
Table 6-2 Summary of results for 8mm tests
Ultimate Strength (MPa)
Modulus of Elasticity (MPa)
Ultimate Elongation (%)
Average Reference Samples 1374 59990 2.30
Average of Cold Exposed Samples
that Ruptured 1370 63540 2.10
Percent of Reference Value 99% 106% 91 %
The standard deviations of the ultimate strength and modulus of elasticity for the cold
exposed tests in Table 6-2 are 2.7% and 9% of their respective averages. By comparing the
averages in Table 6-2, the ultimate strength showed no significant decrease in value as a
result of the cold temperatures, which shows that either matrix microcracking did not occur
or its effect was minimal on the bar properties. The observed small drop in elongation
coupled with the associated increase in stiffness is expected as a result of the cold
temperatures. The 4 samples that failed by debonding of the metal couplers failed at an
63
average tensile stress of 1141 MPa, approximately 250 MPa lower than the failure stress of
the other samples. A photo of a typical debonding failure is shown in Figure 6-1 and a photo
of a typical rupture failure is shown in Figure 6-2. Averaging the measured modulus of
elasticity for all 15 cold exposed samples results in a value of 63536 MPa, which is
approximately 6% higher than the average of the three reference samples.
Figure 6-1 Typical debonding failure
Figure 6-2 Typical rupture failure
64
6.1.2 Results of the tensile test on 12mm bars
The results of all 18 tests conducted are shown in the Table 6-3. In the case of the
12mm samples, only 4 out of the 18 (22%) bar specimens ruptured, the higher frequency of
coupler failure is mainly due to the higher tensile force required to fail the specimen. The
increase in bonded area for the coupler due to the larger diameter is not as great as the
increase in tensile force required to fail the bar. Of particular note is specimen TCB12-04 in
which the coupler steel ruptured before the bar, a photograph of the failed coupler is shown
below in Figure 6-3. Statistically, the result of TCB12-04 is considered to be similar to a
coupler slip failure because the bar did not rupture. Averages calculated from the 3 samples
that ruptured are summarized in table 6-4.
Table 6-3 Test Results for 12mm ComBAR Samples
Ultimate
Strength (MPa)
Modulus of Elasiticity
(MPa)
Ultimate Elongation
(%)
Failure Type Notes/ Remarks
TCB12-01 1168 60087 1.94 Coupler Slip TCB12-02 1089 - - Coupler Slip TCB12-03 1141 53792 2.12 Rupture TCB12-04 1133 59730 1.90 Coupler
Rupture
TCB12-05 1105 53100 2.08 Coupler Slip TCB12-06 1160 54215 2.14 Rupture TCB12-07 1160 - - Coupler Slip TCB12-08 1160 53667 2.16 Coupler Slip TCB12-09 1089 53808 2.02 Coupler Slip TCB12-10 1022 54856 1.86 Coupler Slip TCB12-11 1192 58862 2.02 Rupture TCB12-12 1082 55285 1.96 Coupler Slip TCB12-13 1117 58730 1.90 Coupler Slip TCB12-14 1073 - - Coupler Slip TCB12-15 1105 - - Coupler Slip TCB12-16 1160 56270 2.06 Rupture Control (RT) TCB12-17 1110 63400 1.75 Coupler Slip Control (RT) TCB12-18 1114 60900 1.83 Coupler Slip Control (RT)
65
Figure 6-3 Photograph of ruptured coupler in sample TCB12-04
Table 6-4 Summarized Results of 12mm Tests
Ultimate Strength
(MPa) Modulus of Elasticity
(MPa) Ultimate Elongation
(%) Average of Reference
Samples 1160* 60190 2.06*
Average of Cold Exposed Samples that
Ruptured 1164 55165 2.09
Percent of Reference Value
100 % 92% 99%
* Values use TCB12-16 rupture strength/elongation
For the specimens that ruptured, the data again is consistent with little scatter with the
standard deviation of the ultimate strength and modulus being 2.2% and 5.3% of their
respective averages. For the 12 specimens that failed by coupler debonding the average stress
was 1115 MPa which is about 96% of the average of the ruptured samples. It should be noted
that some of the samples failed by debonding at loads equal to or greater than the load at
which a sample ruptured which seems to indicate that for this group of tests, the capacity of
the coupler is close to the capacity of the bar itself. The average modulus of elasticity based
on all fifteen cold exposed samples is 55893MPa which differs from the reference samples
by about 7%.
6.1.3 Results of tensile tests on 16mm
A total of 17 tests were conducted on 16mm bars, these tests are different from the
other tests on smaller bar sizes in that the majority did not use adhered strain gauges to
measure the elongation. Measurements of the elongation were taken from the head
66
displacement of the loading machine and corrected using the LVDT readings for the initial
reference sample. This calculation assumes a constant amount of slip for each of specimens
and constant slip regardless of temperature which was not the case. Therefore two specimens,
TCB16-14 and 16-15 were instrumented with 320 ohm strain gauges. In addition, specimen
TCB16-16 failed by coupler debonding at a very low load and is omitted from any
subsequent analysis.
Table 6-5 Test Results for 16mm ComBAR Samples
Ultimate
Strength (MPa)
Modulus of Elasiticity
(MPa)
Ultimate Elongation
(%)
Failure Type Notes/ Remarks
TCB 16-01 1236 63902 2.83 Rupture Control (RT) TCB16-02 1119 52246 2.14 Slip TCB16-03 1260 49349 2.55 Slip TCB16-04 1278 52317 2.44 Slip TCB16-05 1271 59676 2.13 Slip TCB16-06 1167 62037 1.88 Slip TCB16-07 1251 56133 2.23 Slip TCB16-08 1203 55034 2.19 Slip TCB16-09 1254 54809 2.29 Rupture TCB16-10 1143 49128 2.33 Rupture TCB16-11 1220 52930 2.31 Rupture TCB16-12 1209 55632 2.17 Rupture TCB16-13 1134 52845 2.15 Rupture TCB16-14 1117 58716 1.90 Rupture Strain Gauged TCB16-15 1254 55268 2.27 Rupture Strain Gauged TCB16-16 1079 - - Slip -
Seven out of the 16 (43.75%) cold exposed samples ruptured while the remaining
ones had a de-bonding failure between the coupler and the bar. Similar to the 12mm bars,
more than 50% slipped primarily due to the much higher tensile force required to fail the bar.
Summarized on the following page are the averages of the ruptured cold temperature samples
compared against the control sample. Because of the scatter in data for the 16mm bars, the
reference test was also compared against 5 test results provided by the manufacturer with
good correlation.
67
Table 6-6 Summarized Results of Select 16mm Tests
Ultimate Strength (MPa)
Modulus of Elasticity (MPa)
Ultimate Elongation (%)
Average of 5 Reference Tests
provided by Manufacturer
1271 64273 1.98
Reference Sample (TCB16-01) 1236 63902 1.93
Average of TCB16-9 to 16-13 (A) 1192 53069 2.25
Average of TCB16-14 & 15 (B) 1186 56992 2.09
Average of all Cold Exposed Samples 1198 54723 2.21
Percent of Reference Value (TCB16-01)
(A) 96% 83% 116%
Percent of Reference Value (TCB16-01)
(B) 96% 89% 107%
Percent of Reference Value (TCB16-01)
(All Specimens) 97% 86% 114%
The results of the 6 cold exposed samples that ruptured were split into two separate
groups to account for different instrumentation setups. Specimens 16-9 to 16-13 (A group)
had elongation measured using the corrected head displacement readings which resulted in a
slight overestimation of the elongation as more slip seemed to occur at the lower
temperatures. The results of 16-14 and 16-15 (B Group) were determined using adhered
strain gauges and thus gave better results when compared against the reference sample which
did not have strain gauges and LVDT measurements were used for analysis. Similar to the
case of the 12mm bars, loads for the bars that failed by coupler slipping were similar to the
failure loads for the ruptured bars. This indicates that the capacity of the coupler was
approximately equal to the capacity of the bar.
68
6.2 Large Beam Tests
6.2.1 Load deflection response Data from all of the LVDTs were used to calculate the midspan deflection and the
results were plotted against the load readings from the load cell in the MTS machine. The
displacement stroke measurements from the MTS had inherent errors and were thus not used.
Plotted in Figure 6-4 are the results from five beams on one plot.
Figure 6-4 Load Deformation Response of All Large Beam Samples
The load deflection responses of the 5 beams exhibit some of the key behaviour
characteristics of a GFRP reinforced section including a bilinear behaviour and linear-elastic
behaviour to failure after concrete cracking. One feature unique to GFRP RC is the degree of
stiffness at large displacements. As shown in Figure 6-4 all of the beams maintained the post-
cracking stiffness until failure with midspan displacements as large as 45 mm.
In general, the load deformation responses of the beams followed two distinct paths.
The two stiffer responses are due to TCB3201 and TCB3202 having the smallest shear spans.
The mode of failure in all samples was the debonding of the 32mm GFRP bar. The post-peak
69
responses of all the beams shown in the figure are all due to the remaining un-ruptured small
steel bars and debonded ComBAR in the reinforcing cage. The small steel bars maintained
structural integrity of the member after the main 32mm GFRP bar debonded.
Tables 6-7 and 6-8 show the load, moment and displacement for each beam at
cracking and ultimate failure conditions, respectively. Note that standard deviations and
averages were not calculated for the cracking loads and displacements because they are
dependent on the load setup which varied by the test.
Table 6-7 Cracking Load, Moment and Midspan Displacement for all samples
Test Number Cracking Load (kN)
Cracking Moment (kNm)
Midspan Displacement at Cracking (mm)
Loading configuration
TCB 3201 209 137 1.27
720mm
TCB 3202 203 133.5 1.04
720mm
TCB 3203 164 116.8 1.15
500mm
TCB 3204 162 135.7 1.19
TCB 3205 165 137.8 1.52
Average - 132.24 - Standard Deviation - 8.79 - -
*Note: Sketches are not to scale
70
Table 6-8 Failure Load, Moment and Midspan Displacement for all samples
Test Number Failure Load (kN)
Failure Moment (kNm)
Midspan Displacement at Failure (mm)
TCB 3201 697 458 45.9 TCB 3202 623 410 39.1 TCB 3203 542 387 40.4 TCB 3204 572 479 45.6 TCB 3205 506 424 37.8 Average - 432 -
Standard Deviation - 37.1 -
The primary GFRP bar in sample TCB3203 seemed to have debonded earlier than all
the other beams as the failure moment was much lower than that of the other four beams. It is
unclear what the reason was for the lower numbers. It should be noted that the high standard
deviation in the failure moment does not necessarily reflect a high variability in the bar
properties because the mode of failure is bond controlled and not tensile rupture of the bar.
The properties of the concrete played an important role in determining the failure load.
6.2.1 Moment‐curvature response
Strain data from the three horizontal LVDTs at midspan were used to calculate the
strain gradient or curvature. The calculated curvatures are plotted against the corresponding
moments in Figure 6-5.
71
Figure 6-5 Moment curvature responses for all 5 beams
In the case of TCB3201, large flexural cracks developed outside of the range of the
LVDTs used to measure the curvature. At a moment of approximately 200 kNm, an abrupt
change in bending stiffness was observed primarily due to large cracks forming near the load
application points which were outside of the LVDT’s measurement range. Consequently,
rotation began to concentrate at that crack location; hence, the curvature values for that
specimen will not be used in further analysis. In general, for the other beams, the responses
were similar but as mentioned previously, debonding occurred much earlier in TCB3203. All
of tests were terminated post bond failure when the midspan deflection and rotation at the
supports became too large for safety. As was the case with the load deflection responses, all
the beams had no distinct softening of the response prior to failure, maintaining stiffness at
very high curvatures.
6.2.2 Bar stress‐strain response
72
Data from the three strain gauges mounted on the bars were used to estimate the
modulus of elasticity of each of the embedded 32mm GFRP bars. By using the longitudinal
strain data, it was possible to locate the neutral axis location, then assuming that the
compression force in the concrete balances the tensile force in one GFRP bar and the 4
smaller steel bars, the GFRP bar stress can be calculated from the applied moment. Because
the mechanical properties of the large GFRP bars were not known prior to testing, one of the
goals of the strain gauges was to help in determining the bar’s modulus of elasticity.
Based on concepts of equilibrium, it was possible to calculate the stress in the
reinforcing bar by making the following assumptions.
1) After cracking, all longitudinal steel bars have yielded. 2) The transverse FRP wrap has no bearing on the flexural strength and calculations. 3) Linear strain distribution at midspan (plane sections assumption).
Shown in Table 6-9 are the calculated bar stresses at failure. Two separate numbers
for each test are reported, one using the method described in the above paragraph and an
additional estimate after incorporating tension stiffening and its effects into the calculation.
Subsequent calculations and bar stress-strain plots were all done incorporating tension
stiffening. Tension stiffening behaviour was modelled using the Vecchio-Collins 1986 model
(Vecchio, 1986).
Table 6-9 Peak Bar Stresses for all Large Beam Tests
Test No. Peak Bar Stress without Tension
Stiffening (MPa) Peak Bar Stress With Tension
Stiffening (MPa) TCB 3201 988 908 TCB 3202 954 892 TCB 3203 833 768 TCB 3204 998 939 TCB 3205 865 809
Calculating the modulus of elasticity requires the use of compatibility relationships.
The measured strains from the gauges mounted on the bars were plotted with the
corresponding calculated bar stresses to produce the stress strain plots for each of the GFRP
bar samples from which the modulus of elasticity could then be determined. Each of the five
73
beams had three gauges resulting in a total of 15 separate stress strain curves. All 15 of the
curves are shown on three separate plots (Figures 6-6 to 6-8). Individual stress-strain plots
from each of the 15 gauges can be found in the appropriate appendix. Note: Peak stress
values on the figures below do not match the data in Table 6-9 as only values obtained from
strain gauges, while they were operating, were used; no extrapolated values are included in
the figures.
Figure 6-6 GFRP bar stress strain plots for TCB3201 & 3202
74
Figure 6-7 GFRP bar stress strain plots for TCB3203
Figure 6-8 GFRP bar stress strain plots for TCB3204 & 3205
75
For each stress-strain curve, three separate modulus of elasticity values were
calculated, two from selecting points around linear portions of the plot and taking the linear
slope, and another from a linear regression of the entire post cracking stress-strain response.
All of the estimates of the modulus of elasticity are shown in the Table 6-10. For the
estimates using linear regression the R2 value is listed to show the error of the estimate.
Table 6-10 Estimates of GFRP Bar Modulus of Elasticity for 5 Large beams
TCB 3201 Gauge Modulus of Elasticity
(MPa) 5-1 55500 5-1 53320 5-1 Regression 55000 (R2=0.999) 5-2 60920 5-2 57080 5-2 Regression 60000 (R2=0.999) 60 59380 60 60890 60 Regression 61000 (R2=0.999)
TCB 3202 Gauge Modulus of Elasticity
(MPa) 5-1 57100 5-1 59250 5-1 Regression 65000 (R2=0.991) 5-2 60550 5-2 60100 5-2 Regression 78000 (R2=0.992) 60 61150 60 62700 60 Regression 64000 (R2=0.984)
TCB 3203 Gauge Modulus of Elasticity
(MPa) 5-1 57100 5-1 59250 5-1 Regression 56300 (R2=0.987) 5-2 65970 5-2 65570 5-2 Regression 55220 (R2=0.969) 60 61150 60 62700 60 Regression 60386 (R2=0.989)
76
TCB 3204 Gauge Modulus of Elasticity
(MPa) 5-1 48500 5-1 52040 5-1 Regression 45000 (R2=0.971) 5-2 51320 5-2 51090 5-2 Regression 43000 (R2=0.983) 60 96775 60 83515 60 Regression 71000 (R2=0.942)
TCB 3205 Gauge Modulus of Elasticity
(MPa) 5-1 50000 5-1 49890 5-1 Regression 40000 (R2=0.985) 5-2 54560 5-2 52630 5-2 Regression 50000 (R2=0.930) 60 76940 60 78150 60 Regression 71000 (R2=0.988)
The influence of shear on the results seems to be significant as the specimens where
gauges were in the shear span seems to have a much larger scatter in their data. Also due to
the variability and scatter in the data, the regression analyses are not all consistent with the
corresponding manually calculated values. Only the manually calculated moduli of elasticity
values are used to estimate the overall modulus of elasticity. From the numbers in Table 6-11,
the average estimate of the modulus of elasticity from all tests is 57200 MPa with a standard
deviation equal to 8.7% of the average; data from the 60mm gauges in the last two tests were
omitted from the modulus of elasticity estimate because some of their results were
considered erroneous. In addition, the length of the 60mm gauges in the 3 point tests were
measuring an average strain over a distance of varying stresses which result in an
overestimation of the stiffness. Testing reported by the manufacturer indicate that the
approximate modulus of elasticity is close to 60,000 for that size bar.
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Table 6-11 Summary of modulus of elasticity estimates
Test No. Gauge No.
Calculated Modulus of Elasticity
Overall Average for Each Test
TCB 3201
5-1 54410 57850 5-2 59000
60 60135 TCB 3202
5-1 58175 60140 5-2 60325
60 61925 TCB 3203
5-1 58180 61960 5-2 65770
60 61925 TCB 3204
5-1 50270 50740 5-2 51205
60 - TCB 3205
5-1 49945 51770
5-2 53595 60 -
6.2.3 Crack width behaviour
Visible cracks were found to be very few in number but very wide as the use of
external glass FRP made any small cracking difficult to see and identify. Because these
specimens were reinforced with one large high strength bar, fewer cracks are to be expected.
The space and width of cracking were also dependent on the loading arrangement in the
beams. In the specimens with larger shear spans, cracks were noted to be spaced farther apart
and wider. Crack patterns on each of the beams are shown in the appendices. The largest
observed flexural crack was 45mm post bond failure. Crack widths at bond failure were not
as wide as those reported in the appendices because cracks began to widen significantly after
the de-bonding failure as the specimens were deformed well beyond their peak.
6.2.4 Bond and stress development behaviour
The de-bonding of the primary load carrying GFRP reinforcing bar was the common
failure mechanism in all five beams. The de-bonding was quite sudden and occurred without
any visible warning, the only indication being a small softening of the load deflection
behaviour before failure.
78
All 5 beams were analyzed to estimate their failure bond stresses. First, the moment
diagram at failure was plotted and then based on those moments along the span, the GFRP
bar stress was estimated. The moment diagrams and bar stresses are plotted in Figures 6-9
through 6-13. When calculating the bar stresses, it is assumed that the smaller #3 bars have
yielded at all locations where the moment exceeds the cracking moment, as well the neutral
axis is assumed to vary linearly from 325mm at locations where the concrete remains
uncracked to the measured height at bond failure at midspan.
Figure 6-9 Bar Stress and Moment Diagram for TCB3201
Analysis Region
Bearing Plate
79
Figure 6-10 Bar Stress and Moment Diagram for TCB3202
Figure 6-11 Bar Stress and Moment Diagram for TCB3203
Analysis Region
Bearing Plate
Analysis Region
Bearing Plate
80
Figure 6-12 Bar Stress and Moment Diagram for TCB3204
Figure 6-13 Bar Stress and Moment Diagram for TCB3205
Analysis Region
Bearing Plate
Analysis Region
Bearing Plate
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The bond stress analysis was done for a region between the cracking moment and
midspan. The region which remains un-cracked (close to the support bearing plate) was
ignored in the bond stress analysis because calculations indicate that the bar stress is minimal.
The analysis region was then subdivided into smaller segments (bond lengths) and the bond
strength was calculated for each segment. Segment sizes ranging from 2.5 bar diameters to
19.5 bar diameters were used.
Table 6-12 Summary of Calculated Bond Strengths
Maximum Bond Strength (MPa)
Minimum Bond Strength (MPa)
Average Bond Strength from all Segments (MPa)
TCB3201 2.5 db (75mm) 6.77 5.45 6.04 5 db (150mm) 6.65 5.50 6.04
7.5 db (250mm) 6.54 5.57 6.06 (425mm) 6.38 5.69 6.04
TCB3202
2.5 db (75mm) 6.10 5.11 5.59 5 db (150mm) 6.05 5.14 5.59
7.5 db (250mm) 5.98 5.20 5.60 (425mm) 5.86 5.30 5.58
TCB3203
2.5 db (75mm) 5.07 4.23 4.65 5 db (150mm) 5.02 4.25 4.62
7.5 db (250mm) 4.95 4.34 4.64 (425mm) 4.87 4.39 4.63
TCB3204
2.5 db (75mm) 5.39 4.14 4.73 5 db (150mm) 5.34 4.16 4.71
7.5 db (250mm) 5.27 4.20 4.75 (725mm) 5.05 4.37 4.71
TCB3205
2.5 db (75mm) 4.71 3.58 4.11 5 db (150mm) 4.65 3.47 4.06
7.5 db (250mm) 4.86 3.57 4.06 (725mm) 4.38 3.68 4.03
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The bond strength estimates in both tables indicate that for this experiment, the
average bond stresses at failure were in the vicinity of 4.1 to 6.0 MPa with the maximum
value of 6.77 MPa. Pull-out tests on smaller sized bars (16mm ComBAR bars) are
summarized in Table 6-14 below (Volkwein, 2007).
Table 6-13 Pull-Out Bond Strengths for 16mm GFRP Bar
Sample
Concrete Cube
Strength
(MPa)
Anchorage
Length (mm)
Peak Tensile
Force (kN)
Average Bond
Strength (Mpa)
B1 69.4 80 (5 db) 68.5 16
B2 71.5 80 (5 db) 67.4 15.8
B3 73.7 80 (5 db) 68.2 16
C1 84.9 80 (5 db) 98.2 23
C2 85.5 80 (5 db) 98.2 23
C3 86.1 80 (5 db) 82.6 19.3
The bond strengths estimated from the beams range between 3.68 and 6.77 with an
average of approximately 5.01 MPa, significantly lower than the bond strengths in the range
of 16 MPa to 23 MPa as reported from pull-out tests. Two major factors can potentially be
the cause of the large difference, the first being the bar size and the second being
confinement. Research into the influence of bar diameter on pull-out bond strength has
indicated that there is a size effect in bond as larger bars displayed lower bond stresses
(Achillides, 2004). Achillides et al. theorize that the influence of shear lag is one of the
reasons for the lower average bond stresses, due to bar stresses at the outside surface being
significantly greater than those in the core of the bar. As described in section 3.2.3, the topic
of shear lag has been disputed as steel bars also display a size effect in bond but are isotropic.
Poisson’s effects have also been shown to adversely affect the bond performance of larger
bars as these bars display larger radial straining and a reduction in cross section, this
shrinking can reduce the ability of the bar to mechanically anchor into the surrounding
concrete (Achillides, 2004).
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The second and more dominant effect for the reduced bond strength in the beam
samples compared against pull-out tests is confinement. The beneficial effects of
confinement have been shown to increase the bond strength by a factor of 3 (Tastani and
Pantazopoulou, 2002). Research into the effects on confinement has shown that for a 13mm
GFRP bar, pull-out tests indicate a bond stress of 20 MPa. Similar bars were tested in 21
specimens where the surrounding concrete was in tension, similar to the beams in this study
and average bond strengths just barely above 5 MPa was reported (Tastani and
Pantazopoulou, 2002). This difference shows that the bond strengths determined through
pull-out testing are not reasonable for design and that the level of confinement plays a pivotal
role in determining the bond strength.
It can be observed from photographs of the failed specimens that the de-bonding
occurred as a failure of the interface between the concrete and GFRP bar. Shown in the
Figure 6-14 is one of the bars inside a crack in the anchorage zone after bond failure. The
damaged rib on the bar in the picture was likely caused by abrasion with a stirrup or tie
during de-bonding.
All of the bars were able to develop at least 830 MPa and in one case reached a stress
of 998 MPa, so no premature mechanical failure of the reinforcing bar was observed. As well
the failure mode was not related to inter-laminar shear of the bar.
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Figure 6-14 Debonded GFRP bar inside large beam post failure
6.3 Prediction of Large Beam Samples
6.3.1 Sectional analysis (Response 2000)
In this section, a simple yet powerful sectional analysis tool, Response 2000, (Bentz
2000) is used to predict the behaviour of GFRP reinforced concrete. Listed in Table 6-14 are
the actual values and parameters used in the analysis. Figure 6-15 shows the experimental
moment curvature plots for specimens alongside the Response 2000 prediction.
Damaged Rib
85
Table 6-14 Summary of parameters in Response 2000 analysis
Concrete Cylinder Strength 69 MPa
Stress-Strain Curve Popovics Curve Compression Softening Model Vecchio-Collins 1986
Tension Stiffening Bentz 1999
GFRP Bars 32 mm Tensile Strength 1000 MPa Elastic Modulus 50000 MPa
Ultimate Elongation 4*% Stress-Strain Response Linear
Steel Bars #3 Yield Strength 400 MPa
Ultimate Strength 600 MPa Elastic Modulus 200000 MPa
Strain for Strain Hardening 0.7% Rupture Strain 10%
* Recommendation to double elongation in Response 2000 technical issue for accurate analysis (Bentz, 2000).
Figure 6-15 Response 2000 moment curvature prediction with experimental results
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The moment curvature response predicted by the computer software has the general
shape of the curves from the test data for TCB3203, 3204 and 3205. One major difference is
that Response 2000 is unable to account for a bond-slip failure as in its computational
algorithms perfect bond is assumed. One cause for the deviation from the prediction in the
high curvature regions is the way the curvatures are calculated for the beam samples. By
using LVDTs, errors can accumulate as the horizontal strains calculated can include the
cracks that form inside the LVDT range. In the case of Beam TCB 3201, the rotation and
cracking primarily occurred outside of the LVDT’s range.
Response 2000 is typically used as a sectional analysis tool however the software
does have the capability to predict member responses. The shear behaviour is modeled in
Response 2000 by solving the 15 simultaneous equations of the Modified Compression Field
Theory (Bentz et al. 2006). One issue with using Response 2000 is that it is difficult to model
the external FRP wrap as shear reinforcement. The external wrap was modeled by decreasing
the metal stirrup spacing to match the capacity provided by the combination of FRP and the
internal steel stirrups. Small effects on shear crack widths are expected by making this
change however, because minimum shear reinforcement is provided in both cases, crack
widths should be well controlled. Results of the Response 2000 member analysis are
presented in Section 6.3.3.
6.3.2 Non linear finite element analysis (VecTor2)
In order to accurately model the bond failure and incorporate bond development into
the prediction, it is necessary to use more advanced software like VecTor 2. This software is
a non-linear finite element program that is part of the VecTor suite for reinforced concrete
analysis; the programs were developed at the University of Toronto by Vecchio and the
VecTor research group (Vecchio, 2008). FormWorks and Augustus, the respective pre and
post processing software were also used to model the beam and interpret the results.
All of the concrete beams were modelled using a mesh of rectangular elements. The
transverse steel stirrup and FRP reinforcement were modelled as smeared reinforcement in
the concrete beam. The longitudinal steel bars were modelled as truss bars with perfect bond
87
to the concrete. Finally, the large 32mm GFRP bar was modelled as a truss element with the
bond between the bar and concrete modelled with contact elements. The model is shown in
Figure 6-16.
Figure 6-16 Mesh for VecTor 2 Analysis
In order to best represent the bond-slip relationship of the GFRP bar, the Eligehausen
model for bond was used with a confinement pressure factor of 1 (VecTor 2 Imperfect Bond).
The bond-slip curve in the model matches well the experimentally determined bond slip
curves was from testing on GFRP bars. Based on the results of that analysis, a second
analysis was done assuming perfect bond (VecTor 2 Perfect Bond). A deflection controlled
analysis was conducted in which the vertical midspan deflection was increased in stages of
0.25mm. The results of the analysis were processed in Augustus. Details on the VecTor 2
structure parameters are provided in appendix C.
6.3.3 Results of analysis procedures
The load deflection response of each of the beams in the study with VecTor 2 and
Response 2000 predictions are presented in Figures 6-9 to 6-11.
88
Figure 6-17 Load deflection of TCB3201 & 3202 with software analysis predictions
89
Figure 6-18 Load deflection of TCB3203 with software analysis predictions
90
Figure 6-19 Load deflection of TCB3204 & 3205 with software analysis predictions
The predictions from Response 2000 for all the beams were generally quite good. The
predictions were able to accurately capture the rapid transition from gross to cracked section
behaviour. The only exception being the prediction of TCB3203 in which the overall
member stiffness was overestimated, this is either due to the earlier than expected bond
failure or measurement errors. Response 2000 is currently only able to model perfect bond
between the concrete and reinforcement which would explain the general under-prediction of
the deflections for all the beams analysed.
In terms of the finite element analysis in VecTor 2, the results had excellent
agreement with the experimental results for the range of loads modelled. Crack patterns and
bar stresses at loads of up to 80% of ultimate failure had good agreement.
The predicted failure mechanism in VecTor 2 for all the beams was sectional shear
failure adjacent to the point of load application. For all the beams with perfect bond assumed,
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the predicted failure loads were between 82 and 97% of the experimental failure load.
Analyses in which the reinforcement is not perfectly bonded resulted in greater crack widths
and resulted in a lower shear failure load. This prediction of a premature shear failure load is
consistent with VecTor 2 models run on other GFRP reinforced beams and slabs. A slab
specimen (# 8F) tested by Nawy et al. (1971) was also modeled and the failure load was
predicted to be 11.6 kN while the experimental failure load was 14.2 kN.
The cause for the low shear strength predictions in the program seems to stem from
an overestimation of the longitudinal strains and crack widths in the member at the
reinforcing bar level. As a result of the premature shear failure in VecTor 2, the peak bond
capacity of the reinforcing bar was never tested. For the analyses conducted with perfect
bond, the analytical and experimental results had very good agreement. A preliminary
investigation on the effect of mesh size on the load deflection behaviour of the beams
indicated that mesh size did not have any significant effect.
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7 DESIGNING WITH GFRP
7.1 Canadian Design Codes for GFRP RC
Currently there are two codes in Canada with provisions for the design of concrete
structures with FRP bars. These are CAN/CSA S6-06 and CAN/CSA S806-02. Shown below
in Table 7-1 presents a summary of some of the major differences in the codes relating to
GFRP design.
Table 7-1 Key differences between S806-02 and S6-06
Code CSA S806-02 CSA S6-06 (CHBDC)
Restriction on Bar Sizes
No Bars Larger than 25mm in Diameter No Restrictions
Material Resistance Factor for GFRP Bars
(φf) 0.75 0.5
Allowable Flexural Failure Mechanisms Concrete Crushing Only Concrete Crushing / Bar
Rupture / Balanced Failure
Modelling in Strut and Tie / (Arch Action) Not Permitted Allowed
Basis of Shear Equations
1994 Canadian Code Simplified Method Equations
Simplified Modified Compression Field Theory
(2004 General Method) with Modifications
Stress Limitations for FRP Bars
30% under Factored Sustained Loads
25% for Full Service Loads (SLS Limit State)
Deformation Performance
Measures None, Crack Limitations only J-Factor
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The large number of differences in key areas of design shows that modifications to
S806-02 are needed. The results from the experiments on large beams reported in chapter 6
indicate that the restriction on bar sizes in S806-02 is not warranted. That clause was
developed based on tests from earlier generation products which may have had interlaminate
shear failure but newer generation products have shown that they are not prone to such
failures (Ehrenstein 2007) as they can develop high tensile stresses.
7.2 International Codes for Design
In addition to the two codes in Canada, there are other codes or guidelines which are
either stand-alone codes or amendments to existing steel reinforced concrete standards. They
are as follows:
• United States of America
• “Guide for the design and construction of concrete reinforced with FRP bars”, American Concrete Institute (ACI) 440.1R-03 2006
• United Kingdom • “Interim Guidelines on the design of reinforced concrete structures using fibre
composite reinforcement”, Institute of Structural Engineers (IstructE), 1999 • Japan
• “Recommendation for design and construction of concrete structures using continuous fibre reinforcing materials”, Japan Society of Civil Engineers (JSCE) 1997
• Italy • “Istruzioni per la Progettazione, l’Esecuzione ed il Controllo di Strutture di
Calcestruzzo Armato con Barre di Materiale Composito Fibrorinforzzato” (CNR-DT 206/2006) National Research Council, Italy 2006
• Norway • NS3473 – Provisional Code on FRP Design
While the differences between these 5 codes and the two Canadian ones would be too
numerous to list, it is important to look at the material resistance factors and stress limitations
for GFRP materials in each of the codes which are summarized in Table 7-2. By quickly
scanning the factors it is clear that there is no general consensus on the durability and
reliability of GFRP materials. Note: Italian code was not included in the comparison.
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Table 7-2 Materials Resistance Factors and Stress Limits from Various Codes for GFRP Design (fib, 2006)
Factor ACI 440.1R-06 NS3473 CHBDC
S6-06 CSA
S806-02 JSCE IstructE
Material Resistance
Factor
Ce 0.7 – Wet 0.8 - Dry
ηenv 0.5
φf 0.5
φf 0.75
1/γfm 0.77
1/γfm 0.77
Reduction for Sustained Stresses/
Permanent Load
Permanent Load Stress
Limits:
0.14-0.16
Conversion Factor: 0.8 – 1
Only in PT applications
Stress Limitation
0.3
0.8 x Endurance
Limit, < 0.7
0.3
Total Strength Reduction for
SLS
0.39-0.52*
*Incorporates φ of 0.55-0.65 depending on
Exposure
0.4-0.5
Stress Limits 0.25
Crack Width
Limitations at SLS only
0.77 0.3
The fib Task Force 9.3 has indicated that no real consensus exists in not only the
resistance factors and stress limits but also the exposure conditions that are considered in the
development of the factors.
The authors of fib Bulletin 40 made note of the fact that the ACI 440 provisions only
consider two different exposure environments, wet and dry, whereas many other factors like
thermal effects can degrade the material. The CHBDC is the only code with no explicit
conditions dealing with sustained stresses for non prestressed applications; however
uncertainty under sustained load is believed to be accounted for in the low material resistance
factor of 0.5. The same can likely be said for the Norwegian code NS3473.
With more and more testing on the durability of FRP bars and internationally
accepted standards being developed, it is likely that the differences in future versions of the
codes will be greatly reduced.
7.3 Proposed Design Methodology for GFRP Bars
Most current GFRP design methodologies and codes consider failure of the GFRP
bars catastrophic and recommend that it should be avoided. While there is no denying the
95
catastrophic nature of that failure, in preventing these failures almost all design
methodologies have departed from traditional limit state design and into a form of the older
working stress design methods.
A design methodology with GFRP based on tensile rupture of the reinforcement is
discussed in this section that is based on limit state design while at the same time maintains
structural safety.
7.3.1 Flexural design
Traditional reinforced concrete design methods rely heavily on the inelastic behaviour
of steel for maintaining structural safety after the design loads were exceeded. GFRP
concrete does not have that ability and thus it can be assumed to behave somewhat more like
a typical prestressed concrete member solely in the regard that rupturing the tensile
reinforcement is entirely possible in some designs.
Two different scenarios are possible in the flexural design of reinforced concrete, the
first being that failure is defined by crushing of the concrete and the other is defined by the
yielding/failure of the reinforcement. In the case of GFRP reinforcing, the latter case is the
more catastrophic. A potential third mode design is a balanced failure design which is rare
and not considered here.
The flexural design method discussed below is solely for the sections controlled by
tensile rupture of the GFRP bars. Over-reinforced sections controlled by concrete crushing
will not be discussed in any detail as their design follows closely a design for a typical over-
reinforced steel section. Sections designed with large compressive regions, like a T beam or
box beam typically have very shallow neutral axis depths which significantly strain the
tensile reinforcement because of the large strain gradient and have a high chance of rupturing
the tensile reinforcement. To design these sections to fail by concrete crushing would require
a significant level of over reinforcing well exceeding the strength requirements. Congestion
of tensile reinforcement would also be a significant problem. Thus the economical solution
96
could be to design for tensile rupture. The following three points should be considered in the
design:
o The number of bars should be selected such that the factored resistance is 1.5
times the factored moment to meet code provisions for tensile controlled
sections.
o In general, a higher total stiffness of the reinforcing cage (EA) provides a
more desirable behaviour at SLS by better controlling crack widths and
deflections.
o Higher strength GFRP bars will result in a higher total failure load of the
reinforcement which provides higher strengths at very large strains and
curvatures.
As is the case with traditional RC design, reinforced concrete stress block analysis
will be used to calculate the moment resistance of the section. Formulations for the stress and
strain in the reinforcement can be developed as well as similar ones for the region of concrete
compression which are shown below:
(7.1)
(7.2)
2 (7.3)
The predefined stress block factors to calculate concrete compression force as given
in the Canadian Code assume a concrete strain of -3.5x10-3 which is not true for sections
controlled by tensile rupture of the reinforcement. In this case, the stress block factors must
be calculated using the following equations.
(7.4)
97
(7.5)
At this point, to solve the equations, iteration is required. A peak concrete strain
needs to be estimated. From equations 7.4 and 7.5, the stress block factors can be determined
and the neutral axis depth (c) has to be iterated until the tensile force from all the bars
balances the compressive force from the concrete. Finally the tensile stress (or strain) in the
bars needs to be compared against the bar rupture stress (strain). If the calculated stress is the
same as the rupture stress the iteration can be stopped, if not, change the top strain value and
repeat until convergence is reached. For sections with multiple layers of reinforcement, each
layer must be analysed individually as the strains differ with depth. The method is shown in a
flow chart in Figure 7-1 below:
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Figure 7-1 Flow Chart for Tensile Rupture Controlled Design Flexural Strength Calculation
After completing the flexural strength design and check, other checks like crack
control and spacing must be made as well. While the process is iterative, all of the equations
are continuous over a range of values and do not require looking up values in tables, a
process that lends itself well to spreadsheet applications. It is likely that additional bars will
need to be provided to satisfy other requirements like crack control and deformability
requirements as the strength requirements do not often govern the design of GFRP reinforced
elements.
99
7.3.2 Designing for Shear with GFRP
Because of the low modulus of elasticity of GFRP bars, the moment shear interaction
is significant in GFRP reinforced members. Designing for the shear strength is of high
importance even in members like slabs. The method for designing for shear is essentially the
same as for steel reinforced members except the separate equations are used for FRP
reinforced members. Currently all equations relating to the shear strength of GFRP
reinforced members in design codes are empirically derived. Recent research has shown
however that the simplified modified compression field theory (SMCFT) equations can be
used with some changes when trying to predict the shear strength of GFRP members (Hoult
et al. 2008).
The issue of shear design becomes far more complicated when providing transverse
shear strength using FRP stirrups. Due to the current fabrication process for making hooked
bars, the strength at bend locations is greatly reduced when compared to a similar straight bar.
In addition, shear design with GFRP transverse reinforcement is typically based on limiting
strains in the stirrup, as very large transverse strains are required to fail a GFRP stirrup. Bent
GFRP bars will not be dealt with in any detail here.
Some manufacturers like Schöck and Pultrall are working on mechanical anchors to
be used with vertical bars as transverse reinforcement. The concept seems to provide the
shear reinforcement but needs to be verified with testing. A setup involving straight anchored
vertical bars will depend heavily on the bond and anchorage characteristics of the bars and
their anchor heads. The adequacy of these specific components is another part of the research
program at the University of Toronto and is not dealt with in this report. Also, externally
applied FRP sheets can be used as transverse reinforcement similar to what was done in the
large beams reported in this study.
7.3.3 Quantifying ductility
Before any method can be presented on performance-based design, measures of
performance must be established. One of the most important measures for steel reinforced
100
concrete is ductility, defined as the ratio between the ultimate and yield
deformations/curvatures. Because of no true yield behaviour, ductility in its traditional sense
does not exist when dealing with GFRP. It is for this reason that many researchers and design
codes have suggested other measures which are collectively referred to as pseudo-ductility
measures.
Bakis et al. (2002) in their paper on GFRP in concrete structures identified that the
major measure of pseudo-ductility is the deformability index. The index is a ratio of the
ultimate deflection and the service deflection; the same can be done with curvatures as well.
This index was based initially on tests conducted by Vijay and Gangarao in 1997 and has
formed the basis for many performance measures. In the following sections, three models
which form the basis of code clauses in both the American and Canadian codes are presented.
7.3.4 Vijay and Gangarao 2001 (DF Factor)
Vijay and Gangarao in 2001 updated their original deformability index to incorporate
strain energy instead of just comparing deformations. As ductility and energy dissipation are
both key concepts in seismic and high performance design, incorporating the two into one
index is useful. They named the term DF for ductility factor.
The DF factor for a beam is determined by comparing the strain energy (area under
the moment curvature diagram) at two levels, one at ultimate failure and the other at some
limiting value of curvature. The limiting curvature value proposed by Vijay and Gangarao is
based upon serviceability indices from ACI 318R99. Based on satisfying typical deflection
and crack width limitations in the code, a limiting curvature of 0.005/d was proposed,
according to the authors this limit corresponds typically to a bar and concrete strain of
4.5x10-3 and -0.5x10-3 respectively. Their conclusions regarding their proposed method were
the following:
• DF factors based on a limiting curvature of 0.005(d) (rad/mm) seem to satisfy
deflection, crack width and energy absorption limits and requirements.
101
• Typical DF factors for over reinforced beams failing via crushing of concrete are in
the range of 6.7-13.9.
• Higher amounts of tensile reinforcement provide higher DF factors.
7.3.5 Yost & Gross 2002 (EFS Design) factor and method
In 2002, Yost and Gross proposed a different pseudo-ductility factor and design
methodology based on a comparison of strain energy density which they name the EFS or
Energy Factor of Safety. While the general idea is the same as the DF factor proposed by
Vijay and Gangarao (2001), the EFS factor is a comparison of material strain energy
densities at ultimate and service conditions instead of the overall member strain energies.
Service conditions are defined as the allowable service moment which would depend on the
design conditions.
The authors then went on to develop a new design methodology based on EFS
principles. By limiting the strain energy densities at the material level (16.5 for FRP bars)
Yost and Gross proposed that their EFS method would provide safe and ductile designs. One
concern is that by limiting the strain energy densities in the materials, designers are reverting
back to a modified form of a working stress design. The authors also stated that working
stress design is appropriate for a brittle elastic material.
An example of designing a bridge according to AASHTO clauses and EFS methodology is
presented in their paper. Some of the key ideas and highlights of the methods are listed
below.
• Minimum reinforcement ratio ρmin = 1.33ρbal
• Allowable concrete service stress = 0.35 f’c
• Strength factor of safety will work out to be approximately 5
• Energy factor of safety will work out to be approximately 28
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7.3.6 CHBDC code J‐Factor and design equations
One of the first implementations of pseudo-ductility principles was in Canada as part
of the CHBDC code requirements for FRP reinforced structures. Jaeger (1995) first proposed
a J-Factor for measuring the deformability and performance of an FRP reinforced concrete
structure. Mufti et al. in 1996 and Jaeger et al. in 1997 further elaborated on the factor and in
2000 it was finally incorporated as a standard in the CHBDC (Bakht et al. 2000).
The J-Factor is based on a comparison of strain energy at ultimate conditions and a
limiting curvature to represent service conditions but does not measure the area under the
moment-curvature curve. It instead measures a rectangular region defined by the origin and a
point of interest at opposite corners. The J-Factor equation as it is found in the CHBDC is as
follows:
(7.6)
J is defined as the overall performance factor which is a ratio of the products of
moment and curvature (Mult ψult) at ultimate and at a limiting condition (Mc and ψc). The
limiting moment and curvature are defined for a maximum concrete compressive strain of -
1x10-3, which is 50% larger than the one recommended by Vijay and Gangarao (2001).
According to code requirements, a minimum J-Factor of 4 is required for rectangular
beams while a minimum factor of 6 is required for T Beams. Those factors are meant to be
very similar to those required of conventionally designed steel reinforced concrete beams
(Bakht et al. 2000).
The CHBDC is unique in that unlike the other Canadian code, it allows GFRP
sections to be designed to fail in tension. CSA S806-02 requires sections to be over
reinforced so that they fail via concrete crushing. Tension failure is allowed in the CHBDC
provided that the section meets all other requirements including the J-Factor as well as an
additional requirement that the factored resistance must be greater than 1.5 times the factored
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moment. Because of the catastrophic nature of tensile rupture, the J-Factor and deformability
of the section become critical part of the design.
In particular, the J-Factor requirement is very strict in the case of a T-Beam. T-Beams
typically fail via yielding/rupture of tensile reinforcement, well before the concrete in
compression reaches the crushing strain required for failure. Hence, the limiting curvature
concrete strain of -1 x 10-3 is not very far from a typical failure strain. What compounds the
issue is that the J-factor requirement for a T-Beam is 50% greater than a rectangular beam.
The J-factor, as well as all of the deformation and performance indices are based on
sectional analysis procedures. Using moment curvature relations to describe the deformation
behaviour makes one significant assumption, that the effects of shear are negligible.
Because of the low stiffness of GFRP bars, the moment shear interaction plays a
significant role in determining the failure mode of the member. Flexurally designed members
without transverse reinforcement will likely ultimately fail in shear and never reach the
flexural failure condition on which the performance measure is based. One other
consideration is that the J-factor requirement also applies to members that are not designed as
flexural members but are designed to carry shear and loads through arching action.
While the J-Factor provides a method to quantify the performance and pseudo-
ductility of a GFRP reinforced member, some aspects of the requirement, namely the T-
Beam requirements need be critically reviewed.
7.4 Design Example of a One-way Slab Reinforced with GFRP Bars
To further illustrate the concept of designing a GFRP RC member to fail by rupture of
reinforcement, a sample one way slab was designed using only GFRP bars as the primary
reinforcement. The slab was proportioned and designed as a flexural member and is similar
to elements like bridge approach slabs. One simplification in the design was that a 4.8 kPa
live load was used in place of a CL-625 truck load. Key design parameters and discussion are
104
presented in this section. For a step by step design with detailed commentary refer to
Appendix D.
7.4.1 Brief Summary of Design
High strength GFRP bars (ComBAR) and normal strength concrete were selected for
this example and design requirements of CSA S6-06 were adopted as CSA S806-02 does not
allow rupture of the reinforcement. The geometry and support conditions of the sample slab
are shown in Figure 7-2 below. A summary of some of the geometry and key material
properties are summarized in Table 7-3.
Figure 7-2 Geometry and Loading of Sample Slab
Table 7-3 Key Design Details for Sample Slab
Geometry and Loading: Span (mm) 6000 Height (mm) 330 Width (mm) Analysed in 1000mm widths Support Conditions Simply Supported Loading Type Uniformly Distributed Factored ULS Moment (kNm) 95.6 (144 for Tensile Rupture) Material Properties: Concrete Cylinder Strength (MPa) 40 GFRP Design Strength (MPa) 1150 Bar Diameter (mm) 16 Bar Area (mm2) 200
The flexural design of the section followed the method detailed in section 7.3.1 for a
member controlled by rupture of reinforcement. When appropriate material resistance factors
were applied, the calculations indicate that the failure is governed by rupture of the GFRP
reinforcement. To meet strength requirements, five bars are required per meter width of slab
(ie: 16M@200mm) when a minimum of 35mm clear cover is used. Further checks indicated
105
that additional bars would be required to meet crack width limitations at service limit states,
thus one additional bar per meter would be needed (ie bar spacing reduces to 170mm). The
shear strength of the slab was found to be adequate and no transverse reinforcement was
needed to meet code requirements. The J-Factor was manually calculated for the section and
was found to be equal to a value of 5.78. Detailed calculations for that J-Factor can be found
in Appendix D.
While the factored moment capacity calculations indicate that failure is governed by
rupture of the reinforcement, when factors are removed the true failure mode in flexure is
concrete crushing. This fact is due to the material resistance factor for FRP being 0.5; having
such a low number indicates that flexural failure can occur at potentially 50% of the true bar
rupture stress. In addition, for rectangular sections like the one designed in the example, bar
rupture is unlikely with all of the additional bars required to meet serviceability and stress
limitations.
7.4.2 Analysis and Discussion of Sample Design
The sample slab was analysed in Response 2000 and the moment curvature plot is
shown below in Figure 7-3. The deflection under service load levels was estimated by
Response 2000 to be approximately 17mm or Ln/320. It should be noted that Response 2000
which uses Modified Compression Field Theory formulations indicate that the ultimate
failure mode of the slab is sectional shear failure, the corresponding moment at shear failure
predicted by Response 2000 is shown on figure 7-3 along with other key points.
106
Figure 7-3 Moment Curvature Response for Sample Slab
Based on the computer analysis values, the J-Factor can be calculated using equation
7.6 and taking values from the figure above. By using the failure moment and curvature (327
kNm, 77 rad/km, respectively) predicted by Response 2000, the following J-Factor can be
calculated.
350 80125 20 11.2
If the moment at shear failure (227 kNm, 44.5 rad/km) is used then the J-Factor
reduces significantly to:
238 43.5125 20 4.14
Both J-Factors meet code requirements for a rectangular section, however, the effects
of shear are significant in determining the overall performance of the section. It should be
Mc for J-Factor
ULS Design Moment (CHDBC) for Bar Rupture
Manually Calculated Failure
Moment
Failure Moment Predicted by
Response 2000
Midspan Moment at Predicted Shear Failure
of Slab
107
noted that mathematically, the J-Factor is calculating the area of rectangle bound by the
points of interest and the origin not the area under the moment curvature diagram like other
measures of pseudo-ductility. For comparison purposes, a similar factor is calculated using
the same points of interest but only the area under the moment curvature curve is considered.
Numerical approximation (Trapezoid) techniques are used to provide a reasonable estimate
of the area without generating any closed form equations. The areas used in the calculations
are summarized in the figure below.
Figure 7-4 Areas for determination of energy dissipation
Based on the areas outlined two factors were calculated. First a conventional factor
based on flexural failure: (Areas: A+B+C / A) is calculated. For the second factor the effects
of shear are included: (Areas: A+B / A). The values are summarized in Table 7-12 below
Mc for J-Factor
Midspan Moment at Predicted Shear Failure
of Slab
A B C
Failure Moment Predicted by
Response 2000
108
Table 7-4 Summary of performance measures for sample slab
Flexural Failure Overall Member Failure
Ratio of Energy Dissipations 9.41 3.43
J-Factor formulations 11.2 4.14
J-Factor based on Code Restrictions
(Assumption of Concrete Strain, etc.)
5.78 -
The J factor calculated based on code provisions (5.78) using code restrictions on the
ultimate concrete strain are conservative estimates of the true energy dissipation behaviour
(11.42). In flexure, the concrete section is controlled by crushing of concrete and Response
2000 does not impose the limit of -3.50 x 10-3 as the crushing strain of concrete as the code
formulations do. This fact leads to higher estimates of failure curvatures. The neutral axis
depths and flexural lever arms agree well between hand calculations and Response 2000 but
the degree of straining in the concrete and the resulting GFRP bar stresses are higher
resulting in higher estimates of the failure moment. Due to the way the J-factor is formulated,
the roughly 20% increase in failure moment predicted by Response 2000 translates into a 50%
increase into the J-Factor. As well, the method of comparing the ratio of the rectangular areas
(11.42) provides reasonable estimates of the true energy dissipation ratio (9.41); however this
method does give slightly unconservative results. Inclusion of the shear effects reduce the
energy factors by 60% from 11.42 to 4.14 because shear failure occurred before flexural
failure. The dependence of the J-Factor on sectional analysis values should be critically
reviewed as the effects of shear and its interaction with moment can be quite significant as
illustrated in this example.
7.4.3 Moment‐Shear Interaction
Because of the low axial stiffness of GFRP, for a given moment, the degree of
longitudinal straining in a GFRP member is much greater than a similarly designed steel
member. This increased degree of longitudinal straining has been a significant factor in
determining the shear strength of GFRP reinforced members (Hoult et al. 2008). It is for this
reason that when designing with GFRP, the moment and its influence on the shear strength is
109
of importance. The moment shear interaction plot generated by Response 2000 for the
sample slab is shown in Figure 7-5.
Figure 7-5 Moment Shear Interaction for Sample Slab
As shown in the figure above, the shear strength of the slab drops off significantly once
the member is subjected to a significant moment. The reason is due to the increased crack
widths which reduce the ability of shear forces to be transmitted along cracks via aggregate
interlock. It should be noted that each point on the envelope is a combination of shear and
moment that occur at the same location in the member. This interaction is one of the reasons
why when flexural failure is difficult to achieve in members without transverse
reinforcement like the example slab. The moment shear interaction is significant largely due
to the low axial stiffness of GFRP bars and should be taken into consideration when
designing any GFRP reinforced member, particularly those without transverse reinforcement.
It should be noted that this concept is taken into consideration in shear provisions based on
SMCFT when calculating εx, however codes not using SMCFT based provisions like CSA
S806-02 do not consider this effect.
Unsafe
Safe
J-Factor Flexural Failure Moment
110
7.5 Hybrid Section Design
The lack of inelastic behaviour and low curvature stiffness in members reinforced
entirely with GFRP flexural reinforcement is a well documented weakness of the material.
Combining layers of steel and GFRP reinforcement in a hybrid section can balance the
strengths of both materials and result in a section that behaves well structurally and is also
very durable.
7.5.1 Principle of hybrid design
Corrosion of reinforcement is a function of three things, the susceptibility of the
reinforcement, the exposure environment and time itself. Two of the key factors in
determining the susceptibility of the reinforcement to corrosion are the material type and its
location. Changing the material into GFRP removes the susceptibility of the reinforcement
significantly reducing the risk of corrosion. Increasing the cover to the reinforcement can
also reduce the susceptibility and in turn reduce the risk of severe corrosion.
That principle forms the basis of the hybrid section design. Placing GFRP as the
outermost layer of reinforcing bars in a member increases significantly the cover depth to the
first layer of susceptible steel reinforcement. Also placing GFRP bars as the outer layer
makes better use of the higher capacity of the GFRP reinforcing bars. Current design
practices often use epoxy coated steel at the outermost layers and black steel for the layers
farther away from exposed surfaces. The principle described in this section is the same,
except that GFRP can be used at the outermost surface layer.
7.5.2 Comparative study of reinforcement types and layouts
Shown below is an analysis of a typical T-beam section in which varying
arrangements of reinforcement types are used and the moment curvature behavior of the
section for all the arrangements is compared. A T-beam was chosen because of its high
utilization of tensile reinforcement. Figure 7-6 shows details of the section and material
properties.
111
Concrete Strength:
• 50 MPa (Popovics Stress Strain
Curve)
Reinforcing Types:
• 20M - 400 MPa Epoxy Coated Steel
• 20M – GFRP Rebar
(1150 MPa ComBAR)
Modelling Software
• Response 2000
Figure 7-6 Concrete Section used for Hybrid Section Analysis
Shown in Figure 7-7 is a comparison of the moment curvature responses for four
different beam sections that are defined in Table 7-5:
Table 7-5 Key for section names
Section Reinforcement Details
All Epoxy Steel All 3 Tensile Layers are Epoxy Coated Steel
Bottom Layer ComBAR Outermost Layer is GFRP, Other 2 Layers
are Epoxy Coated Steel
Bottom 2 Layers ComBAR Outermost 2 Layers are GFRP, Other Layer
is Epoxy Coated Steel
All ComBAR All Tensile Layers are GFRP
4
112
Figure 7-7 Moment Curvature Responses for all 4 Sections
Incorporating layers of GFRP into the design significantly increases the moment
capacity of the section but at the same time reduces the stiffness of the member at low
curvature. The section reinforced entirely with GFRP reinforcing bars has the highest
moment capacity (almost 3 times the steel reinforced one) but also has the softest response at
low curvatures. The other two hybrid sections as shown in the graph fall somewhere in
between sacrificing some of the high moment capacity of a GFRP reinforced section for the
increased stiffness of a steel reinforced section. Another key point regarding the case in
which only the bottom layer is GFRP relates to the failure mode; after the GFRP layer has
exhausted its capacity the two remaining steel layers maintain section integrity at large
deformations. The moment resistance of that section at large curvature values is nearly the
same as the original yield capacity of the all steel case because of the strain hardening that is
occurring in the two steel layers.
For most service load conditions and designs, the low curvature region is the area of
importance because curvatures well beyond typical steel yielding are seldom seen except for
extreme loading conditions. The low curvature region of Figure 7-7 is expanded and shown
in Figure 7-8.
113
Figure 7-8 Enlarged Low Curvature Region of Moment Curvature Responses
Using only one layer of GFRP seems to have a minimal decrease of stiffness at the
low curvature while at the same time nearly doubling the moment capacity of the section.
The benefits also extend beyond the strength concepts as the outer reinforcing layers are now
corrosion resistant. The depth to the first “susceptible” layer is now nearly doubled because
of the GFRP layers, a depth that for most situations is beyond typical levels of chloride
penetration. Another indirect benefit to using hybrid sections rather than all GFRP-reinforced
is the increased longitudinal stiffness of the reinforcement as a whole which will beneficially
increase the shear strength of the section by reducing mid depth longitudinal strains. The
beneficial increase in the shear strength was not covered in the analysis.
Hybrid sections as shown in the above figures have the potential to greatly improve
both the structural and durability performance of structures. These conclusions however, are
drawn based on a simple analytical study only. They should be the subject of experimental
testing and evaluation before being fully utilized in structures. It should also be noted that no
real provisions or guidance exists in any CSA code for a hybrid design.
114
7.6 Summary on the Design of GFRP-Reinforced Concrete Members
In this chapter, the available design provisions for GFRP RC were discussed. The
provisions of the two codes in Canada for designing with GFRP were found to differ greatly
in some very key areas including the material resistance factor, restriction on bar sizes and
allowable analysis techniques. When comparing the material resistance factors and stress
limits from codes around the world for FRP, it was also found that no real consensus exists
on the design of GFRP-reinforced members for strength or durability.
In the particular case of GFRP reinforced concrete section design to fail by bar
rupture, a design method was summarized and illustrated with a design example. In addition,
because of the low modulus of elasticity of GFRP bars, the shear strength becomes a design
concern in all member types including slabs. Several available pseudo-ductility measures
were also discussed and compared. It was found that all of the different measures omit the
influence of shear and its interaction with the moment capacity whose influence was found to
be significant in the design example. All performance measures including the J-Factor in the
new S6-06 are based on sectional analysis techniques and assume that failure in shear does
not occur.
115
8 CONCLUSIONS
8.1 General Conclusions on GFRP bars and GFRP Reinforced Concrete
From a review of the available literature on currently available GFRP bars, it was
shown that the strengths and properties of GFRP bars varied considerably from one
manufacturer to other. Available data shows that GFRP bars are reasonably durable under a
variety of exposure conditions including strong alkali solutions, sustained loading and
extreme heat. Results from field tests on 10 year old structures indicate that in many cases
the simulated lab experiments over-estimate the degree to which degradation occurs on
GFRP rebars in concrete structures.
Tests conducted on three sizes of GFRP bars have shown that extreme cold
temperatures do not have a significant effect on the bar mechanical properties. Matrix
microcracking either did not occur or did not have an appreciable effect on the bar properties.
Five beams reinforced with large 32mm high strength GFRP bars were also tested under
monotonic three- and four- point bending. It was determined that bond and anchorage played
a pivotal role in the structural response and that development lengths required to fully
develop the large GFRP bar were in excess of 1.5m (greater than 45 bar diameters). Data
from surface mounted strain gauges also indicated that the modulus of elasticity of the large
32mm GFRP bars was on average 57,200 MPa. The 32mm GFRP bars were able to develop
stresses ranging from 806 to 998 MPa without any interlaminar shear failure, the ultimate
failure mechanism in all 5 samples being bond failure at the concrete and bar interface.
Both sectional analysis and non-linear finite element software analysis programs
(Response 2000 and VecTor 2) were both shown to predict the experimental results of the
beam tests well for the range of loads analyzed. In the case of VecTor 2, the analysis
terminated prematurely due to a predicted sectional shear failure. Prior to termination of the
analyses, the results of the VecTor 2 analysis were very close to the experimental values.
116
Based on the results presented in this paper and previous work from other researchers,
it can be concluded that using GFRP reinforcement does not change the fundamental flexural
behaviour of reinforced concrete. Current analysis techniques work well in predicting the
response of GFRP RC members like stress block analysis for flexure. A sample slab using
16mm GFRP bars was designed and then analyzed in Response 2000. Results of the analysis
show that while the factored moment capacity indicate the slab is controlled by rupture of
reinforcement, the flexural failure mode is actually governed by concrete crushing. In
addition, while the section met deformability requirements (J-Factor), the final mode of
failure was shear and when the interaction of shear forces is included in the analysis, the
overall deformation performance decreases significantly, a behaviour not accounted for in
the code provisions.
8.2 Future work
While the results from cold temperature testing indicate that there is no significant
drop in properties from the reference samples tested, the database is still small especially for
large size bars.
For the large GFRP bars, because of their dependence on the bond development of
reinforcement, an in-depth study into the bond behaviour of the newer high strength bars is
urgently needed. Also, to compensate for the higher bond demand and unreliable bend
strength some manufacturers are making mechanical anchors for their bars. The behaviour
of these anchors needs to be investigated as having reliable bond strengths could open many
other possibilities of using GFRP including tension ties in beams and vertically anchored
shear reinforcement for slabs and beams.
The sample design presented in Chapter 7 indicates that further investigation into the
effects of shear on the overall design of GFRP RC is lacking. While the design of the section
in Chapter 7 satisfied major design criteria like stress limits and crack widths, the slabs’s
ultimate failure mode is shear which is undesirable. Such an investigation could provide
refinement to measures of performance like the J-Factor in the current CHBDC. With
117
difficulties in providing transverse shear reinforcement, an investigation into the behaviour
of mechanically anchored bars as shear reinforcement should be conducted.
From a preliminary analysis, the idea of reinforcing sections with a combination of
GFRP bars and steel bars have shown promise in being able to balance the strengths and
weaknesses of GFRP and steel while providing a durable system. These sections could be
further explored in a more in-depth study.
118
9 REFERENCES
Test and Technical Reports
Colberg, R., (2007) “Tensile Testing of ASLAN 100”, Report RB4_SO_6469, July 2007
Colberg, R., (2007) “Tensile Testing of ASLAN 100”, Report RB5_SO_7034, Aug 2007
Colberg, R., (2007) “Tensile Testing of ASLAN 100”, Report RB6_SO_7108, Aug 2007
Colberg, R., (2007) “Tensile Testing of ASLAN 100”, Report RB7_SO_6361, July 2007
Dejke, V., (2002) “Durability and Service Life Prediction of GFRP for Concrete
Reinforcement”, Chalmers University of Technology. 2002.
Ehrenstein, G.W. (2007), “Expert Report – GFRP – Reinforcing Bars “ComBAR””
Report No: STN Report Erlangen 2007 01 EN, 2007.
Ehrenstein, G.W. (2007), “Expert’s Comments on the GFRP Reinforcement ComBAR”,
University of Erlangen, October 2007.
El-Gamal, S., & Benmokrane. B., (2008) “Tensile properties of #5 V-Rod-HM Glass FRP
Bars”, Technical Report University of Sherbrooke, May 2008.
Kiefer, D., (2006) “Monitoring of the tensile tests performed on ComBAR GFRP
reinforcement”
Report No: 06 14 50 0446, 2006.
Kreuser, K. (2008), “Dauerschwingversuche an einbetonierten Bewehrungsstaben Schöck
ComBAR mit 16mm Durchmesser.”, Report No: 08 24 16 0256, 2008.
119
Kreuser, K. (2007), “Dynamic fatigue tests of Schöck ComBAR reinforcing bars, with a
diameter of 16mm, in concrete.”, Report No: 06 24 15 0662-1, 2007.
Kreuser, K. (2007), “Dynamic fatigue tests of Schöck ComBAR reinforcing bars, with a
diameter of 16mm, in concrete. Upper stress = 300N/mm2”, Report No: 06 24 15 0662-2,
2007.
Nause, I. (2005), “Determination of temperature-dependant tensile strengths of ComBAR
reinforcement bars.” Report No: 072/05-Nau-3740/6345, Brunswick Institute for Concrete
Material Testing. Germany. June 2005.
Pultrall Inc. (2007), “Technical Data Sheet V-Rod”, Pultrall Inc., Thetford Mines, Que.,
Canada, March 2007.
Schmachtenberg, E., & Trawiel, P. (2007), “Test Report AP-07-11-169-1 /tr”, University of
Erlangen, December 2007.
Schmachtenberg, E., & Trawiel, P. (2008), “Test Report AP-08-02-37 /tr”, University of
Erlangen, Germany, March
Sheikh, S.A. & Johnson, D.T.C. (2008), “Cold Temperature Tensile Testing of Schöck
ComBAR”, Research Report SJ07-01, Department of Civil Engineering, University of
Toronto, December 2008.
Shiekh, S.A. & Johnson, D.T.C. (2007), “Compliance of 16mm ComBAR GFRP Bars with
ISIS Canada Certification Specifications”, Department of Civil Engineering, University of
Toronto, October 2007.
Volkwein, A. (2007), “Pull-out Tests”, Report submitted to Schöck Bauteile GmbH, 02,
2007.
120
Weber, A. (2007), “Durability tests performed on straight ComBAR GFRP bars with
standard coating d=16mm”. Report No: 116 05 G, Schöck Research Labs, Baden Baden,
Germany, 2007.
Weber, A. (2007), “16mm Tensile Test Report 25.05.2007”, Report No: 075 07, Schöck
Research Labs, Baden Baden, Germany, 2007.
Weber, A. (2007), “8mm Tensile Test Report 25.05.2007”, Report No: 075 07 ComBAR-8, ,
Schöck Research Labs, Baden Baden, Germany, 2007.
Weber, A. (2007), “12mm Tensile Test Report 25.05.2007”, Report No: 075 07 ComBAR-
12, , Schöck Research Labs, Baden Baden, Germany, 2007.
Weber, A. (2007), “Expert Report, Application for DIBt Certification of GFRP Reinforcing
Bars “ComBAR” made by Schöck Bauteile GmbH”, Schöck Research Labs, Baden Baden,
Germany, 2007.
Weber, A. (2008), “Vertical bond tests with ComBAR 8mm and 12mm in different concrete
qualities”, Test Report Schöck Bauteile GmbH, 3rd version, February 2008, Schöck Research
Labs, Baden Baden, Germany.
Journal and Conference Papers
Abdalla, H.A. (2002), “Evaluation of deflection in concrete members reinforced with fibre
reinforced polymer (FRP) bars”, Composite Structures, Vol. 56, pp 62-71, 2002
Achillides, Z., Pilakoutas, K. (2004), “Bond Behavior of Fiber Reinforced Polymer Bars
under Direct Pullout Conditions”, ASCE Journal of Composites for Construction, Vol. 8, No.
2, March-April 2004.
121
Al-Salloum, Y., Sayed, S., Almusallam, T. (1997), “Behavior of Concrete Beams Doubly
Reinforced by GFRP bars”, Non-Metallic (FRP) Reinforcement for Concrete Structures,
Proceedings of the Third International Symposium, V.2, pp 471-478, 1997
Al-Salloum, Y., Sayed, S., Almusallam, T. (1996), “Some Design Considerations for
Concrete Beams Reinforced by GFRP Bars”, First International Conference on Composite
Infrastructure (ICCI-96), pp 318-331, 1996
American Society of Civil Engineers (ASCE) (2003), “2003 Progress Report on America’s
Infrastructure” , American Society of Civil Engineers, Reston, Virginia, USA,2003
Bakht, B., Al-Bazi, G., Banthia, N., Cheung, M., Erki, M., Faoro, M., Machida, A., Mufti, A.,
Neale, K., Tadros, G. (2000), “Canadian Bridge Design Code Provisions for Fiber
Reinforced Structures”, ASCE Journal of Composites for Construction, Vol. 4, No. 1,
February 2000.
Bakis, C.E., Bank, L.C., Brown, V.L., Cosenza, E., Davalos, J.F., Lesko, J.J., Machida, A.,
Rizkalla, S., Triantafillou, T. (2002), “Fiber-Reinforced Polymer Composites for
Construction-State-of-the-Art Review”, ASCE Journal of Composites for Construction, Vol
6, No. 2, May 2002
Barris, C., Torres, L., Turon, A., Baena, M., Mias, C. (2008), “Experimental study of the
flexural Behaviour of GFRP Reinforced Concrete Beams” Fourth International Conference
on Composites in Civil Engineering (CICE 2008), Zurich, Switzerland, 2008
Bentz, E.C. (2000), “Response 2000”, http://www.ecf.utoronto.ca/~bentz/r2k.htm, 2000
Bentz, E.C. (2000), “Response 2000 Bugs”, http://www.ecf.utoronto.ca/~bentz/r2k.htm,
2000
122
Benmokrane B., Challal, O., Masmoudi, R. (1996), “Flexural Response of Concrete Beams
Reinforced with FRP Reinforcing Bars”, ACI Structural Journal, Vol. 93, No. 1, January
1996.
Brown, V.L., Bartholomew, C.L. (1993), “FRP Reinforcing Bars in Reinforced Concrete
Members”, ACI Structural Journal, Vol. 90, No. 1, January-February 1993.
Brown, V.L., Bartholomew, C.L. (1996), “Long-term Deflections of GFRP Reinforced
Concrete Beams”, First International Conference on Composite Infrastructure (ICCI-96),
1996, pp 389-400
Chen, Y., Davalos, J.F., Ray, I., Kim, H. (2007), “Accelerated Aging Tests for evaluations of
durability performance of FRP reinforcing bars for concrete structures”, Composite
Structures, Vol 78, 2007, pp 101-111
Federal Highway Administration FHWA (2007), “2006 Status of the Nation’s Highways,
Bridges and Transit: Conditions and Performance”, Report Submitted to Congress, received
from FHWA, January 2007.
Federal Highway Administation FHWA (2007), “Tables of frequently requested National
Bridge Inventory (NBI) Data 2007, Received from FHWA, December 2007.
Gangarao, H.V.S., Faza, S.S., (1991) “Bending and Bond Behavior and Design of Concrete
Beams Reinforced with Fiber Reinforced Plastic Rebars”, Constructed Facilities Center
Report, CFC -92-142, 1991.
Guibe, C., Francilette, J. (1996), “Time-Temperature-Transformation (TTT) Cure Diagrams:
Relationships Between Tg, Cure Temperature, and time for DGEBA/DETA Systems”,
Journal of Applied Polymer Science, Vol 62., 1996, pp 1941-1951.
123
Halliwell, S. (2002), “Fibre reinforced polymer (FRP) composite bars for new build”, NCN
case review on exploiting the benefits of composites, 2002, National Composites Network,
Cambridge, United Kingdom, 2002.
Hoult, N.A., Sherwood, E.G., Bentz,E.C., Collins, M.P. (2008), “Does the use of FRP
Reinforcement Change the One-Way Shear Behavior of Reinforced Concrete Slabs?”, Vol.
12, No. 2, March-April 2008.
Hulder, G. (2008), “Zur Aushärtung kalthärtender Reaktionsharzsysteme für tragende
Anwendungen im Bauwesen”, Dissertation Doktor-Ingenieur (German), University of
Erlangen, 2008.
Jaeger, L.G., Mufti, A.A., Tadros, G., (1997). “The Concept of the Overall Performance
Factor in Rectangular-section Reinforced Concrete Beams.” Proceedings: Third International
Symposium on Non-metallic (FRP) Reinforcement for Concrete Structures. JCI. Vol. 2, pp
551-558
Jaeger, L.G., Tadros, G., Mufti, A.A., (1995) “Balanced Section, Ductility and Deformability
in Concrete with FRP Reinforcement” Technical Report No. 2-1995, Technical University of
Nova Scotia, Halifax.
Katz, A., Berman, N., Bank, L.C. (1999), “Effect of High Temperature on Bond Strength of
FRP Rebars”, ASCE Journal of Composites for Construction, Vol. 3, No. 2, May 1999
Malvar, L.J. (1995), “Tensile and Bond Properties of GFRP Reinforcing Bars”, ACI
Structural Journal, Vol. 92, No. 3, May-June 1995
Masmoudi, R., Theriault, M., Benmokrane, B. (1998), “Flexural Behavior of Concrete
Beams Reinforced with Deformed Fiber Reinforced Plastic Reinforcing Rods”, ACI
Structural Journal, Vol. 95, No. 6, November-December 1998.
124
Michaluk, C., Rizkalla, S., Tadros, C., and Benmokrane, B. (1998), “Flexural behaviour of
one-way slabs reinforced by fiber plastic reinforcement.” ACI Structural Journal, Vol. 95, No.
3, pp 353-365, 1998.
Ministry of Transportation of Ontario (MTO) (2009), “Quick Facts”, obtained from
http://www.mto.gov.on.ca/english/about/quickfacts.shtml.
Mosley, C., Tureyan, A., Frosch, R. (2008), “Bond Strength of Nonmetallic Reinforcing
Bars”, ACI Structural Journal, Vol. 105, No. 5, September-October 2008.
Mufti, A., Banthia, N., Benmokrane, B., Boulfiza, M., Newhook, J. (2007), “Durability of
GFRP Composite Rods”, Concrete International, February 2007
Mufti, A.A., Newhook, J.P., Tadros, G., (1996) “Deformability Versus Ductility in Concrete
Beams With FRP Reinforcement. Proceedings: 2nd International Conference on Advanced
Composite Materials in Bridges and Structures, ACMBS-2, pp 189-199.
Nam, J.H., Lee, S., Yoon, S.J., Jang, W.S., Cho, S.K. (2006), “Flexural Behavior of Concrete
Beam Reinforced with Steel and FRP Re-Bars”, Key Engineering Materials, Vols 306-308,
2006, pp 1367-1372.
Nawy, E., Neuwerth, G. (1977), “Fiberglass Reinforced Concrete Slabs and Beams”, ASCE
Journal of the Structural Division, Vol. 103, No. ST2, February 1977.
Nawy, E., Neuwerth, G., Phillips, C. (1971), “Behavior of Fiber Glass Reinforced Concrete
Beams”, ASCE Journal of the Structural Division, Vol 97, No. ST9, September 1971.
Nkurunziza, G., Benmokrane, B., Debaiky, A.S., Masmoudi, R. (2005), “Effects of Sustained
Load and Environment on Long-Term Tensile Properties of Glass Fiber-Reinforced Polymer
Reinforcing Bars”, ACI Structural Journal, Vol. 102, No. 4., July-August 2005.
125
Pecce, M., Mafredi, G., Cosenza, E. (2000), “Experimental Response and Code Models of
GFRP RC Beams in Bending”, ASCE Journal of Composites for Construction, Vol. 4, No. 4,
November 2000.
Robert, M., Cousin, P., Benmokrane B. (2009), “Behaviour of GFRP Reinforcing Bars
Subjected to Extreme Temperatures”, CSCE Annual Conference, May 2009.
Russell, H.G. (2004), “Concrete Bridge Deck Performance”, National Cooperative Highway
Research Program (NCHRP) Synthesis 333, Transportation Research Board Publication
NCHRP 20-05, November 2004.
Sonobe, Y., Fukuyama, H., Okamoto, T., Kani, N., Kimura, K., Kobayashi, K., Masuda, Y.,
Matsuzaki, Y., Mochhizuki, S., Nagasaka, T., Shimizu, A., Tanano, H., Tanagaki, M.,
Teshigawara, M. (1997), “Design Guidelines of FRP Reinforced Concrete Building
Structures”, ASCE Journal of Composites for Construction, Vol. 1, No. 3, August 1997
Tam, S., Sheikh, S.A. (2008), “Behavior of fibre reinforced polymer (FRP) and FRP bond
under freeze thaw cycles and sustained load”, Fourth International Conference on
Composites in Civil Engineering (CICE 2008), Zurich, Switzerland, 2008
Tang, B.E. (1997). Fibre Reinforced Polymer Composites Applications in USA. First Korea
USA Road Workshop Proceedings January 28-29. FHWA.
Tastani, S.P., Pantazopoulou, S.J. (2006), “Bond of GFRP Bars in Concrete: Experimental
Study and Analytical Interpretation”, ASCE Journal of Composites for Construction, Vol. 10,
No. 5, September-October 2006
Tastani, S.P., Pantazopoulou, S.J. (2002), “Experimental Evaluation of the Direct Tension-
Pullout Bond Test”, Bond in Concrete – From Research to Standards, Budapest, 2002
126
Theriault, M., Benmokrane, B. (1998), “Effects of FRP Reinforcement Ratio and Concrete
Strength on Flexural Behavior of Concrete Beams”, Journal of Composites for Construction
Vol.2, No. 1 February 1998.
Toutanji, H.A., Saafi M. (2000), “Flexural Behavior of Concrete Beams Reinforced with
Glass Fiber-Reinforced Polymer (GFRP) Bars”, ACI Structural Journal, Vol. 97, No. 5.,
September-October 2000.
Vecchio, F.J. (2008), “FormsWorks, VecTor 2 and Augustus Bundle Ver. 2.8”,
http://www.civ.utoronto.ca/vector/software.html, accessed Sept. 2008.
Vecchio, F.J., Collins, M.P. (1986), “The Modified Compression Field Theory for
Reinforced Concrete Elements Subjected to Shear”, ACI Structural Journal, Vol. 83, No. 2,
pp 219-231, March 1986
Vijay, P.V., Gangarao, H.V.S. (2006), “Unified Limit State Approach Using Deformability
Factors in Concrete Beams Reinforced with GFRP Bars,” ASCE Structural Congress
Proceedings, November 2006.
Vijay, P.V., Gangarao, H.V.S. (2001), “Bending Behavior and Deformability of Glass Fiber-
Reinforced Polymer Reinforced Concrete Members”, ACI Structural Journal, Vol. 98, No. 6.,
November-December 2001.
Wambeke, B., Shield, C. (2006), “Development Length of Glass Fiber-Reinforced Polymer
Bars in Concrete. ACI Structural Journal, Vol. 103, No. 1, January-February 2006.
Wang, H. (2004), “Discussion of Bond Behavior of Fiber Reinforced Polymer Bars Under
Direct Pullout Conditions”, ASCE Journal of Composites for Construction, Vol 8, No., 2,
March-April 2004
127
Weber, A. (2008), “Fire-Resistance tests on Composite Rebars”, Paper 2.D.1., Fourth
International Conference on Composites in Civil Engineering (CICE2008), Zurich,
Switzerland, 2008.
Wines J.C., Hoff, G.C. (1966), “Laboratory Investigation of Plastic-Glass Fiber
Reinforcement for Reinforced and Prestressed Concrete Report 1.” Army Engineer
Waterways Experiment Station, Accession Number: AD0630600. Feb.1966.
Yost, J.R., Goodspeed, P.E., Schmeckpeper, E.R. (2001), “Flexural Performance of Concrete
Beams Reinforced with FRP Grids” ASCE Journal of Composites for Construction, Vol. 5,
No. 1, February 2001.
Yost, J.R., Gross, S.P., Dinehart, D.W. (2003), “Effective Moment of Inertia for Glass Fiber-
Reinforced Polymer- Reinforced Concrete Beams”, ACI Structural Journal, Vol. 100, No. 6,
November-December 2003.
Yost, J.R., Gross, S.P., Dinehart, D.W. (2003), “Flexural Stiffness of High Strength Concrete
Beams Reinforced with GFRP Bars”, ACI Special Publication, SP-210-10, February 2003.
Zhao, W., Pilakoutas, K., and Waldron, P. (1997), “FRP Reinforced Concrete: Cracking
Behavior and Determination”, Non-Metallic (FRP) Reinforcement for Concrete Structures,
Proceedings of the Third International Symposium, pp 439-446. 1997
Codes and Standards
American Concrete Institute (2004), “Guide Test Methods for Fiber-Reinforced Polymers
(FRPs_ for Reinforcing of Strengthening Concrete Structures”, ACI 440.3R-04, American
Concrete Institute, Farmington Hills, Mi., 2004
128
American Concrete Institute (2006), “Guide for the design and construction of concrete
reinforced with FRP bars”, ACI 440.1R-06, American Concrete Institute, Farmington Hills,
Mi., 2006
Canadian Standards Association (2004), “Concrete Design Handbook”, CSA A23.3 2004,
CSA Mississauga 2004.
Canadian Standards Association (2006), “Canadian Highways Bridge Design Code
(CHBDC)”, CSA S6-06, CSA Mississauga 2006.
Canadian Standards Association (2002), “Design and Construction of Building Components
with Fibre Reinforced Polymers”, CSA S806-02 2002, CSA Mississauga 2002.
fib (2006), “FRP reinforcement in RC structures”, Technical report prepared by a working
party of fib Task Group 9.3 bulletin 40, International Federation for Structural Concrete,
2006
Intelligent Sensing for Innovative Structures (ISIS) Canada (2006), “Specifications for
Product Certification of FRPs as Internal Reinforcement for Concrete Structures”, ISIS
Canada, Winnipeg Manitoba, September 2006.
IstructE (Institution of Structural Engineers) (1999), “Interim Guidance on the design of
reinforced concrete structures using fibre composite reinforcement”, SETO Ltd., London,
UK, 1999
JSCE (1997), “Recommendation for design and construction of concrete structures using
continuous fibre reinforcing materials”, Research Committee on Continuous Fiber
Reinforcing Materials, Japan, Society of Civil Engineers, Tokyo, Japan, 1997
129
Appendix A
STRESS‐STRAIN PLOTS FOR COLD TEMPERATURE SAMPLES
130
The stress strain responses from all the cold samples tested are presented in this appendix on
individual plots. Dashed lines are extrapolated values. For 16mm samples in which no
gauges were used, strain was determined by the machine movement calibrated by LVDT
readings.
Modulus of elasticity estimates were calculated from the linear elastic portions of the plots,
the failure elongation was estimated by dividing the measured ultimate failure stress by the
calculated modulus of elasticity.
Plots for the following bars are not included in this appendix because the gauges mounted on
the specimen did not function at cold temperatures, or the test failed prematurely.
TCB8-12
TCB12-2
TCB12-7
TCB12-14
TCB12-15
TCB16-16
131
132
133
134
135
136
137
138
139
140
141
142
143
144
145
146
147
148
149
150
\
151
152
153
Note: 16-14 was strain gauged, only region where gauge is operational is plotted.
154
Note: 16-15 was strain gauged, only region where gauge is operational is plotted.
155
Appendix B
BEAM PHOTOS AND DIAGRAMS
156
Photo 1 - North side of TCB32-01
Photo 2 – South side of TCB32-01
157
Photo 3 - North side of TCB32-02
Photo 4 - South side of TCB32-02
158
Photo 5 - North side of TCB32-03
Photo 6 - South side of TCB32-03
159
Photo 7 - North side of TCB32-04
Photo 8 - South side of TCB32-03
160
Photo 9 - North side of TCB32-05
Photo 10 - South side of TCB32-05
161
Appendix C
STRUCTURE DATA FOR VECTOR 2 ANALYSIS
162
Presented below is the structure data used in the finite element model. Note: only the model
for 3 point bending is shown, for the other two models locations of the bearing plates need to
be moved to match the experiment. Element Indicies are omitted.
* * * * * * * * * * * * * * * * * * *
* V e c T o r 2 *
* S T R U C T U R E D A T A *
* * * * * * * * * * * * * * * * * * *
STRUCTURAL PARAMETERS
*********************
Structure Title (30 char. max.) : MASc Project
Structure File Name ( 8 char. max.) : Beam
No. of R.C. Material Types : 2
No. of Steel Material Types : 2
No. of Bond Material Types : 1
No. of Rectangular Elements : 948
No. of Quadrilateral Elements : 0
No. of Triangular Elements : 0
No. of Truss Bar Elements : 216
No. of Linkage Elements : 0
No. of Contact Elements : 72
No. of Joints : 1110
No. of Restraints : 3
163
MATERIAL SPECIFICATIONS
***********************
(A) REINFORCED CONCRETE
-----------------------
<NOTE:> TO BE USED IN RECTANGULAR AND TRIANGULAR ELEMENTS ONLY
CONCRETE
------------------------
MAT Ns T f'c f't Ec e0 Mu Cc Agg Dens Kc Sx Sy
TYP # mm MPa MPa MPa me /C mm kg/m3 mm2/s mm Mm
1 4 400 67.5 0 0 0 0 0 0 0 0 0 0
2 0 400 400 0 200000 0 0 0 0 0 0 0 0
REINFORCEMENT COMPONENTS
------------------------
MAT REF
DIR As Db Fy Fu Es Esh esh Cs Dep
TYP TYP deg % mm MPa MPa MPa MPa me /C me
1 1 90 0.50 9 450 620 200000 10000 2.250 0.000 0.000
1 1 361 0.50 9 450 620 200000 10000 2.250 0.000 0.000
1 5 361 1.3 5.2 450 450 18080 0 25 0 0
1 5 90 1.3 5.2 450 450 18080 0 25 0 0
164
(B) STEEL
---------
<NOTE:> TO BE USED FOR TRUSS ELEMENTS ONLY
MAT REF AREA Db Fy Fu Es Esh esh Cs Dep
TYP TYP mm2 mm MPa MPa MPa MPa me /C Me
1 1 804.5 32 1000 1000 50000 0 20 0 0
2 1 142.5 9.525 450 620 200000 36000 2.250 0 0
165
Appendix D
SAMPLE DESIGN OF RUPTURE CONTROLLED GFRP RC SLAB
166
Simple One-Way Slab – CHBDC Chapter 16 Design Provisions
Note: This appendix is not a design aid and should not be used for the design of GFRP reinforced concrete
members. It is an illustrative tool compiled as part of an M.A.Sc. project and should not be used for actual
design purposes. Refer to relevant code provisions when designing GFRP sections. RESPONSE 2000 is also
not a design tool for the actual design of GFRP reinforced members.
Uniform Loading for the Sample Design
SLS 1 ULS 1 Factored Dead Load (kPa) 8.09 9.70 Factored Live Load (kPa) 4.32 8.16
Factored Super Imp. Dead Load (kPa) 2.12 3.17
Total Factored Load (kPa) 14.53 21.03
Shear and moment along member span
Dv 0.1 L 0.2 L 0.3 L 0.4 L Mid Span L (m) 0.26 0.60 1.20 1.80 2.40 3.00
SLS Moment (kNm) 10.77 23.52 41.82 54.89 62.73 65.34 SLS Shear (kN) 39.81 34.85 26.14 17.42 8.71 0.00
ULS Moment (kNm) 15.60 34.08 60.58 79.51 90.87 94.66 ULS Shear (kN) 57.67 50.48 37.86 25.24 12.62 0.00
Material Properties
Concrete Properties Cylinder Str. (Mpa) 40.00
E (Mpa) 28400 Peak Strain (x10-3) -2.00 Concrete Density
(Kg/m3) 2450.00 Concrete Stress Strain Parabolic
GFRP Bars Diameter (mm) 16.00
Modulus of Elasticity (Mpa) 60000.00 Design Strength
(Mpa) 1150.00 Design
Elongation (%) 1.92
Slab Geometry
Height (h) 330mm Cover to Reinf. 35mm Depth to Reinf. 287mm
Width Analysis Done in
per m widths Clear Span 6000mm
167
Preliminary Flexural Design (Design for Tensile Rupture of Reinforcement)
1. Determine Effective Depth to Reinforcement:
d h cover 12 d
d 330 35 12 16
d 287mm
2. Try 5 GFRP bars /m of slab, (16@200mm)
A FRP 5 200 1000mm
3. Guess Top Concrete Strain (Concrete Crushing: – 3.5 x 10-3)
1st Try: -1.5 x 10-3
4. Calculate Stress Block Factors:
β4
εε′
6 2εε′
4— 1.5
2.06 2 1.5
2.00.72
α1β
εε′
13εε′
1
0.721.52.0
13
1.52.0
0.78
5. Guess Neutral Axis Depth:
1st Try: 50mm
6. Calculate Stresses and Strains at Reinforcement Depths:
εε
1.5 1050
287 50 7.11 10
7.11 10 60,000 426.6
7. Calculate Factored Tensile Force:
0.5 426.6 200 42.7
8. Calculate Total Tensile Force:
213.5 5 213.3
9. Determine Depth of Compression Block for Axial Force Equilibrium
168
′ 213.3
213.3 1040 1000 0.78 0.7
9.8
10. Determine and Verify Neutral Axis Depth
9.80.72
12.6 50
Note: The calculated neutral axis does not match with the original estimate, repeat steps 5 to 10 with
a new estimate of the neutral axis.
2nd Estimate: 26mm (Calculated Values for Both estimate summarized in table below)
C (Guess) (x10-3) fFRP(Mpa) FFRP T a c (Calc)
1st Estimate 50 7.11 426.6 42.67 213.3 9.780952 12.55826
2nd Estimate 26.2 14.9313 895.87786 89.5 447.9389 20.54041 26.3
Now have to Check the Bar Strain: 14.9 19.2. Bars have not ruptured;
estimate of top concrete strain was too small. Repeat steps 3 – 10 with a higher estimate of the top
concrete strain.
2nd Estimate of using spreadsheet solving routines: -1.95 x 10-3
New Values: = 0.75 , = 0.88
c (Guess) (x10-3) fFRP(Mpa) FFRP T a c (Calc)
1st Estimate 26.5 19.17 1150.1 115 575.1 23.3 26.47
11. The Bars have Ruptured with this estimate of can now proceed to calculate the moment capacity:
2575.1 287
23.32
158 /
1.5 158 1.5 94.6
Note: 1.5 Mf is the requirement for sections failing via rupture of GFRP Bars
169
Therefore: Only 5 Bars needed to meet strength requirements.
Serviceability Limit State Check (Crack widths and Bar Stresses)
65.4 /
1. Determine the bar stresses, neutral axis depth at SLS conditions using a method similar to the
Preliminary Flexural Design. For SLS conditions, all load and resistance factors are removed.
Solving with a spreadsheet:
= -0.44 x 10-3, = 0.68 , = 0.3
C (Guess) (x10-3) fFRP(Mpa) FFRP T a c (Calc)
1st Estimate 28.9 3.92 236 47 236 19.65 28.9
2236 287
28.92
65.3 /
2. Calculate the crack width and compare against the specified limit:
h1 330 28.9 301.1
h2 12
301.1 8 35 258.1
fFRP Stress in the FRP bars at Service Loads 236
EFRP Modulus of Elasticity of the Longitudinal Reinforcement 60000
dc Minimum Cover 35
kb Coefficient depending on bond between FRP and Concrete 0.8
s Spacing of shear or tensile reinforcement (mm) 200
2 2 . ..0.8 35 0.5 200 0.77 0.7
The section is not satisfactory and additional bars need to be provided. Add one additional Bar per m
width. New Arrangement : 16M@167mm.
170
2nd Attempt with 16M@167mm
h1 330 28.9 301.1
h2 12
301.1 8 35 258.1
fFRP Stress in the FRP bars at Service Loads 196
EFRP Modulus of Elasticity of the Longitudinal Reinforcement 60000
dc Minimum Cover 35
kb Coefficient depending on bond between FRP and Concrete 0.8
s Spacing of shear or tensile reinforcement (mm) 167
212
219660000
258.1301.1
0.8 35 0.5 167 0.55
0.7
Now the SLS check is acceptable with 6 bars per meter width, the factored moment capacity should
be recalculated for the new amount of reinforcement.
3. Compare SLS Reinforcement Stresses with Limit
SLS bar stress: 236 MPa, 0.17 0.25
Deformation and Performance Factor (J-Factor)
Note: For this calculation all material resistance factors are removed to give an accurate estimate of
the actual performance of the section.
Note: For this calculation the Moment and Curvature for a top concrete strain of -1.0 as well as at
failure are required. Method of calculation is similar to the Preliminary Flexural Design.
171
′ x 10-3 -2.00x-03
x 10-3 -1.00x-03
0.70
α 0.60
fFRP (MPa) 381
T (kN) 457
c (mm) 39.24
Mr (kNm/m) 125
(rad/mm) 2.56 x 10-5
ε′ x 10-3 -2.00x-03
ε x 10-3 -3.50x-03 *
β 0.90
α 0.81
fFRP (MPa) 906
T (kN) 1087
c (mm) 53
Mr (kNm/m) 288
(rad/mm) 6.48 x 10-5
* When material resistance factors are removed the actual section fails by concrete crushing, only
when looking at the factored tensile resistance do the calculations indicate that the section fails via
rupture of reinforcement.
J286 6.48125 2.56
5.78
The J-Factor for this section passes the requirement for a rectangular section, it should be noted that
the section in reality will fail via concrete crushing and not tensile rupture.
Bond and Development Length Calculation
d 0.45k k
d K EFRPES
ff
A
K1 Bar Location Factor 1.0
K4 Bar Surface Factor 0.8
EFRP Modulus of Elasticity of FRP Bar (MPa) 60,000
ES Modulus of Elasticity of Steel (MPa) 200,000
dcs Minimum cover to reinforcement (mm) 35
Ktr Transverse reinforcement factor 0
ffrpu Specified Strength of FRP (MPa) 1150
172
fcr Cracking strength of concrete (MPa) 2.09
A Cross Sectional Area of GFRP Bar (mm2) 200
d 0.450.8 1.035
11502.09
200 1131 mm
Therefore, need to provide at least 1131mm of development length for a single bar or provide some
mechanical anchorage.
Shear Strength Calculation
EFRP Modulus of Elasticity of Longitudinal Reinforcement 60000
h Overall height of section 330
d Effective Depth to Reinforcement 287
dlong = 0.9 d or 0.72 h Effective Shear depth to Reinforcement (Approximation of flexural lever arm) 258.3
bw Web width 1000
Af Total area of longitudinal reinforcement on the flexural
tension side 1200
ag Maximum Aggregate Size (mm) 20
Vc 2.5βϕ f b dEE
ε
Md V V 0.5N A f
2 E A EFRPAFRP` 0.003
β1
1 1500ε1300
1000 s
Similar to the CSA A23.3 General Method of Shear Design, the process of determining the shear
failure load of the slab is iterative.
Ex: at Location 0.1 L, MULS = 34.08 kNm/m, VULS = 50.48 kN/m, M/V = 0.675
173
To converge on a solution, the failure shear load has to be estimated and corresponding moment has
to be calculated, based on those two values, an actual failure shear load can be calculated, the two
will converge on the correct solution.
MMV
V 0.675 V
V (kN/m) M (kNm/m) ε β Vc (kN/m)
90 60.75 0.000625 0.213 122.5
110 74.25 0.000765 0.192 110
The solution converged on 110 kN/m which is greater than the factored ultimate shear load of 50.5
kN/m. The slab does not need transverse reinforcement. Note: Should also check all other key
locations, 0.2L, 0.3L, 0.4L, midspan and distance dlong from supports, they were not included in the
example as 0.1L was the critical location.
174
Behaviour of the Designed Section
Moment Curvature plot for the final design (RESPONSE 2000)
SLS Design Moment
ULS Design Moment
Tensile Rupture Design Moment
Mc for J‐Factor
Ultimate Failure Moment
(Controlled by Concrete Crushing)
Factored Moment Capacity