CHAPTER 8

Stresses within a Soil Mass

v

Part II dealt with the forces that act between individual

soil particles. In an actual soil it obviously is impossibleto keep track of the forces at each individual contactpoint. Rather, it is necessary to use the concept of stress.

This chapter introduces the concept of stress as itapplies to soils, discusses the stresses that exist within asoil as a result of its own weight and as a result of appliedforces, and, finally, presents some useful geometricrepresentations for the state of stress at a point withina soil.

8.

1 CONCEPT OF STRESS FOR APARTICULATE SYSTEM

Figure 8.1a shows a hypothetical small measuringdevice (Element A) buried within a mass of soil. Weimagine that this measuring device has been installed in

such a way that no soil particles have been moved. Thesketches in Fig. 8.1/>,c depict the horizontal and verticalfaces of Element A, with soil particles pushing againstthese faces. These particles generally exert both a nor-mal force and a shear force on these faces. If each face

is square, with dimension a on each side, then we candefine the stresses acting upon the device as

vv-

o' uh - ., > 'n - o v ~ 2 v 1Ja

~

a" a a

where Nv and Nhi respectively, represent the normalforces in the vertical and horizontal directions; Tv andTh, respectively, represent the shear forces in the verticaland horizontal directions; and ov, ah, rv, and rh representthe corresponding stresses. Thus we have defined fourstresses which can, at least theoretically, be readilyvisualized and measured.

//WA\m//A\VA\v/A\V/WA\VAV/A\V/WA Ground surface

(b)

- Element A (c)

(a)

Fig. 8.1 Sketches for definition of stress, (a) Soil profile, (b) and (c) Forces atelement A.

97

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98 PART III DRY SOIL

Cross sectionsthrough particles

Pore space

Point of contact betweenparticles lying below

and above plane

Fig. 8.2 Definition of stress in a particulate system.Z7\.

a x a a x a a x a

In Part III, except as noted, it will be assumed thatthe pressure within the pore phase of the soil is zero;i.

e., equal to atmospheric pressure. Hence the forces Nv,

Nh, Tv, and Th arise entirely from force that is beingtransmitted through the mineral skeleton. In dry soil,stress may be thought ofas theforce in the mineral skeletonper unit area of soil.

Actually, it is quite difficult to measure accurately thestresses within a soil mass, primarily because the presenceof a stress gage disrupts the stress field that would other-wise exist if the gage were not present. Hamilton (1960)discusses soil stress gages and the problems associatedwith them. In order to make our definition of stress

apply independently of a stress gage, we pass an imagin-ary plane through soil, as shown in Fig. 8.2. This planewill pass in part through mineral matter and in partthrough pore space. It can even happen that this planepasses through one or more contact points bet\yeenparticles. At each point where this plane passes throughmineral matter, the force transmitted through the mineralskeleton can be broken up into components normal andtangential to the plane. The tangential components canfurther be resolved into components lying along a pairof coordinate axes. These various components aredepicted in Fig. 8.2. The summation over the plane ofthe normal components of all forces, divided by the areaof the plane, is the normal stress a acting upon the plane.Similarly, the summation over the plane of the tangentialcomponents in, say, the x-direction, divided by the areaof this plane, is the shear stress tx in the x-direction.

There is still another picture which is often used whendefining stress. One imagines a "wavy" plane which iswarped just enough so that it passes through mineralmatter only at points of contact between particles. Stress

is then the sum of the contact forces divided by the areaof the wavy plane. The summation of all the contact areaswill be a very small portion of the total area of the plane,certainly less than 1 %. Thus stress as defined in thissection is numerically much different frora the stress atthe points of contact.

In this book, when we use the word "stress" we meanthe macroscopic stress; i.e., force/total area, the stressthat we have just defined with the aid of Figs. 8.1 and 8.2.When we have occasion to talk about the stresses at the

contacts between particles,' we shall use a qualifyingphrase such as "contact stresses". As wks discussed inChapter 5, the contact stresses between soil particles willbe very large (on the order of 100,000 psi). The macro-scopic stress as defined in this chapter will typically rangefrom 1 to 1000 psi for most actual problems.

The concept of stress is closely associated with theconcept of a continuum. Thus when we speak of thestress acting at a point, we envision the forces againstthe sides of an infinitesimally small cube which is com-posed of some homogeneous material. At first sight wemay therefore wonder whether it makes sense to applythe concept of stress to a particulate system such as. oil.However, the concept of stress as applied to soil is nomore abstract than the same concept applied to metals.A metal is actually composed of many small crystals, andon the submicroscopic scale the magnitude of the forcesbetween crystals varies randomly from crystal to crystal.For any material, the inside of the "infinitesimally smallcube" is thus only statistically homogeneous. In a senseall matter is particulate, and it is meaningful to talk aboutmacroscopic stress only if this stress varies little overdistances which are of the order of magnitude of the sizeof the largest particle. When we talk about the stresses

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at a "point" within a soil, we often must envision a ratherlarge "point".

Returning to Fig. 8.1, we note that the forces Nv, etc.,are the sums of the normal and tangential components ofthe forces at each contact point between the soil particlesand the faces. The smaller the size of the particle, thegreater the number of contact points with a face ofdimension a. Thus for a given value of macroscopicstress, a decreasing particle size means a smaller forceper contact. For example, Table 8.1 gives typical valuesfor the force per contact for different values of stress anddifferent particle sizes (see Marsal, 1963).

Table 8.1 Typical Values for Average Contact Forceswithin Granular Soils c

Particle Particle

Average Contact Force (lb)for

Macroscopic Stress (psi)De- Diameter

scription '(mm) 1 10 100

60 3 30 300Gravel

2.

0 0.

003 0.

03 0.

3Sand

0.

06 3 x 10-6 3 x lO"5 0.

0003Silt

0.

002 3 x 10-9 3 x 10-8 3 x 10-7

8.

2 GEOSTATIC STRESSES

Stresses within soil are caused by the external loadsapplied to the soil and by the weight of the soil. Thepattern of stresses caused by applied loads is usually quitecomplicated. The pattern of stresses caused by the soil'sown weight can also be complicated. However, there isone common situation in which the weight of soil givesrise to a very simple pattern of stresses: when the groundsurface is horizontal and when the nature of the soil variesbut little in the horizontal directions. This situation

frequently exists, especially in sedimentary soils. In sucha situation, the stresses are called geostatic stresses.

Vertical Geostatic Stress

In the situation just described, there are no shearstresses upon vertical and horizontal planes within thesoil. Hence the vertical geostatic stress at any depth canbe computed simply by considering the weight of soilabove that depth.

Thus, if the unit weight of the soil is constant with

Ch. 8 Stresses within a Soil Mass

Stress (au and Oh)

99

Fig. 8.3 Geostatic stresses in soil with horizontal surface.

depth,Gr = zy (8.2)

where z is the depth and y is the total unit weight of thesoil. In this case, the vertical stress will vary linearly withdepth, as shown in Fig. 8.3. A typical unit weight for adry soil is 100 pcf. Using this unit weight, Eq. 8.2 can beconverted into the useful set of formulas listed in Table8.

2.

1

Table 8.2 Formulas for Computing Vertical GeostaticStress

Units for av Units for z Formula for cv

psf feet 1002psi feet 0.6942kg/cm2 meters 0.1582atmospheres feet 0.04732

Note. Based upon y = 100 pcf. For any other unitweight, multiply by y/100.

Of course the unit weight is seldom constant with depth.Usually a soil will become denser with depth because ofthe compression caused by the geostatic stress. If theunit weight of the soil varies continuously with depth,the vertical stress can be evaluated by means of theintegral

Or = y dz (8.3)

1 A complete+ist of factors for converting one set of units to othersis given in the Appendix.

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100 PART III DRY SOIL

If the soil is stratified and the unit weight is different foreach stratum, then the vertical stress can conveniently becomputed by means of the summation

7r = 2 y A2 (8.4)

Example 8.1 illustrates the computation of verticalgeostatic stress for a case in which the unit weight is afunction of the geostatic stress.

Example 8.1

Given. The relationship between vertical stress and unitweight

V = 95 + 0.0007 av

where y is in pcf and av is in psf.Find. The vertical stress at a depth of 100 feet for a

geostatic stress condition.Solution Using Calculus. From Eq. 8.3:

(95 + 0.0007(7,,) dz (z in feet)or

day= 95 + 0

.0007 ov

The solution of this dilTerential equation is:

ar = 135,800((?0 0007 - 1)For z = 100 ft:

gv = 135,800(1.0725 - 1) = 9840 psf

A hernative Approximate Solution by Trial and Error.First trial: assume average unit weight from z = 0 to

z = 100 ft is 100 psf. Then ov at z = 100 ft would be 10,000psf. Actual unit weight at that depth would be 102 pcf, andaverage unit weight (assuming linear variation of y withdepth) would be 98.5 pcf.

Second trial: assume average unit weight of 98.5 pcf. Thenat 2 = 100 ft, ov = 9850 psf and y = 111.9 pcf. Average unitweight is 98.45 pcf, which is practically the same as for theprevious trial.

The slight discrepancy between the two answers occursbecause the unit weight actually does not quite vary linearlywith depth as assumed in the second solution. The dis-crepancy can be larger when y is more sensitive to ov. Thesolution using calculus is more accurate, but the user easilycan make mistakes regarding units. The accuracy of the trialsolution can be improved by breaking the 100-ft depth intolayers and assuming a uniform variation of unit weightthrough each layer. Ph. (d) Decrease of vertical strain. During unloading, the elasticenergy stored in the particles caused upward movement of A relative to B. Reverse sliding must develop tomaintain the condition of zero lateral displacement. Hence Pv < Ph.

Whitman (1963) has discussed the importance of timeeffects during loadings of very short duration.

Small Increments of Stress Superimposed on an InitialStress

The stress-strain behavior is shown in Fig. 10.10.Sliding between particles does not begin until the stressincrement exceeds some critical level. For smaller incre-

ments, strains result only from elastic deformations ofindividual particles (Whitman, Miller, and Moore, 1964).

The stress required to initiate interparticle slidingincreases with increasing initial stress and decreasingvoid ratio. This critical stress is increased when the soil

has been heavily prestressed by previous loadings, and is

larger for rapid loadings than for slow loadings. Formost engineering problems this critical stress is probablyless than 1 psi and hence is of no practical concern.However, this initial range of stress-strain behavior isimportant to the study of wave propagation velocities.

Lateral Stresses during Confined CompressionDuring confined compression, particle motions are,

on the average, in one direction only. Thus when thetangential contact forces are summed over the manycontacts lying on some surface, there should be a nettangential force; i.e., a net shear stress on the surface.Hence, in general, the horizontal stress will differ fromthe vertical stress during confined compression. The

Ch. 10 General Aspects of Stress-Strain Behavior 127ratio of horizontal to vertical stress is, by definition, Kq,the lateral stress ratio at rest.

When a granular soil is loaded for the first time, thefrictional forces at the contacts act in such a directionthat ah is less than ov; i.e., KQ < 1. The magnitude ofK0 must depend on the amount of frictional resistancemobilized at contact points between particles. Figure10.11 shows data giving values of K0 as a function of thefriction angle (f).1 For a few soils, such as the SangamonRiver $and, the value of Kq can be predicted by a theoret-ical equation based upon the study of an idealized packingof elastic spheres. However, experimental values of Kqare best represented by an expression suggested by Jaky(1944):

0=l-sin (10.1)Combining Eq. 10.1 with Eq. 8.12, which defines theslope /9 of the Kq stress path, leads to

and

tan =

sin =

sin 2 - sin

2 tan /S1 + tan p

(10.2)

(10.3)

As indicated in Fig. 10.7d, the direction of the frictionalforces at contact points between particles begins toreverse during unloading. For a given vertical stress, thehorizontal stress will be larger during unloading thanduring the original loading. During the later stages ofunloading, the horizontal stress may even exceed thevertical stress. This pattern is shown by the experimentaldata given in Fig. 10.12. During reloading of a soil, the

1 4) wil! be defined in Chapter 11; it is the peak friction angle.

20

40

60

80

100

Time (mm)

Fig. 10.8 Time curve for a typical load increment on sand(From Taylor, 1948).

lateral stress ratio generally starts out at a value greaterthan that given by Eq. 10.1, and then decreases to thisvalue as the stress increases. During cyclic loading andunloading, the stress path will be as shown in Fig. \0.1b,with the lateral stress ratio alternating approximatelybetween Kq and 1 /KQ.

10.4 STRESS-STRAIN BEHAVIOR DURINGTRIAXIAL COMPRESSION

Figure 10.13 shows a typical set of data from a triaxialtest upon a sand. The stress path for this test is givenin Fig. 10.14. The specimen was first compressed iso-tropically to 1 kg/cm2 by increasing the chamber pressure.Then the vertical (axial) stress was increased while thehorizontal stress (chamber pressure) was held constant.

50

.

1 60

70

80

90

1001 10 100 1000

Time (min)

Fig. 10.9 Typical compression-time curve for high-stress test (FromRoberts, 1964).

128 PART 111 DRY SOIL

Vertical stress

Strains result fromelastic distortions

of particles

Interparticle\ sliding begins

\" \> \

\\\\\\

Fig. 10.10 Behavior during small stress increment super-imposed upon an initial stress.

Figure 10.13 plots q, equal to one-half the deviatorstress, versus the vertical (axial) strain. This stress-strainrelation becomes curved at very small strains and achievesa peak at a strain of about 3 %. The resistance of the soilthen gradually decreases until this test was arbitrarilystopped at a strain of 11.6%. If the test had been carriedto larger strains, the stress-strain curve would haveleveled off at a constant value of stress. For further dis-cussion of this stress-strain behavior, it is useful to definethree stages in the straining process:

1. An initial stage during which strains are very small.

For the test shown in Fig. 10.13 this range extendsto a strain of about 1%.

2. A range which begins when the specimen begins to

yield and which includes the peak of the curve andthe gradual decrease of resistance past the peak.

JAKYKorrl-sin~ 0.5

0.3

o Minnesota sand

x Pennsylvania sandSangamon River sand

A Wabash River sand

0.

3 0.4 0.5

sin 0.

8

5 9 13 17 21 25 29Overconsolidation ratio (O.C.R.)

(b)

33 37

Fig. 10.11 Coefficient of lateral stress at rest for initial loadingversus friction angle (from Hendron, 1963).

Fig. 10.12 Lateral stress during one-dimensional compres-sion. Minnesota sand; e0 = 0.62, Dr - 0.34. (FromHendron, 1963.)

For the test shown, this range extends frorr\strain until

,

the end of the test.3

. A final range during which the resistance is constantwith further straining. ' This range is called theultimate condition.

Behavior During Initial Range

During the initial range the volume of the specimendecreases slightly, as shown in Fig. 10.13. Part (c) of thefigure shows that the specimen is bulging slightly so thatthe horizontal strain is negative, but numerically thehorizontal strain is less than the vertical strain.

This is exactly the pattern of behavior that would beexpected when the compressive stresses are increasing.In this stage the particles' are being pushed into a denserarrangement. The general behavior is very similar to thatduring confined or isotropic compression. Figure 10.15compares the stress-strain behavior during isotropic,confined, and triaxial compression upon identical speci-mens which initially had the same void ratio and carriedthe same vertical stress.

Behavior at and near Peak

Within this range the soil fails. The deviator stress atthe peak point of the stress-strain curve is called thecompressive strength of the soil. The value of q at thepeak (i.e., half the compressive strength) is related tothe shear strength of the soil.

The behavior during this stage is quite different from

Ch. 10 General Aspects of Stress-Strain Behavior 1292

.

0

1.

6

E 1.

2

0.

8

0.

4

p

Ch amber pre ssure = 1*ig/crr 2

I

\

-c ) , o < >-

-o

o

4 6 8

Vertical strain, ev (%)(c)

10 12

Fig. 10.13 Results of triaxial compression test upon a wellgraded calcareous sand from Libya. -

that during the initial stage, and can be explained bystudying the deformation of a planar array of rigidspheres. Figure \0A6d shows a unit element from adensely packed array. When this element is compressedvertically, strains can result only if the spheres C and Dmove laterally. This pattern of motion must be accom-panied by an increase in the volume of the array, as canbe seen by comparing the void spaces in parts (a) and (b)of the figure. Figure 10.136 shows that just such a volumeincrease occurs during loading of actual soils. It is aremarkable fact that a dense sand, wherf compressed in

2.

0

&. 1.

0

- ..

...

_

.

.

1

Lp-yn*Pe

-- -

._

_

-

-

1

7-

-

i1'.

1.

0 2.0

- |Ap (kg/cm2)3

.

0

Fig. 10.14 Stress path for standard triaxial test on wellgraded sand from Libya.

one direction, actually increases in volume. The fact wasfirst observed and investigated by Osbourne Reynolds in1885. Reynolds applied the name dilatant to this volumeincrease effect.

The planar array of spheres can also be used to studythe conditions that exist at the peak of the stress-straincurve and to explain the decrease in strength past thepeak (Rowe, 1962). However, these aspects of thebehavior can be more readily discussed using the sketchesin Fig. 10.17, which illustrate the concept of interlocking.

Figure 10.170 shows soil particles sliding over a smoothsurface. This is the situation we dealt with in Chapter 6,and for this situation the shear resistance is given bythe mineral-to-mineral friction angle. However, thesituation within actual soils is more like that shown inparts (b) and (c) of the figure: soil particles are in contactwith other soil particles, and planes through the contactpoints are inclined to the horizontal. In order to have a

>

Isotropiccompression Confined

compression

Triaxialcompression

Vertical strain

Fig. 10.15 Comparison of stress-strain curves for threetypes of compression.

130 PART III DRY SOIL

3

specimen expands in volume to a very marked degree asthe specimen is strained. On the other hand, the loosespecimen first decreases in volume, then expands onceagain, and finally ends up at almost the volume at whichit started.

150

ST 120

I 90

.K' 60

Q 30

-20

-15

-10

a)_

EE -5

+5Q

0.

9

0.

8

0.

7

0.

6

eo = 0.605

/-

= 0.

834

.

Ch. 10 General Aspects of Stress-Strain Behavior 131The following patterns of behavior are just what would

be predicted by the concepts of dilatancy and inter-locking:

1. The denser the sand, the greater the interlocking,

hence the greater the deviator stress, hence thegreater the friction angle.

2. The denser the sand, the greater the volume increase

which must occur.

3

e ) = 0.605

A,

eo = 0 834

en = (1834

0.

605

10 15 20Axial strain (%)

25 30

p

6

Strain {%)

Fig. 10.18 Stress-strain curves for loose and dense specimens.Medium-fine sand. 03 = 30 lb/in.2:

132 PART III DRY SOIL

4

E2

-1

-2

-3

--

--

i 2

i

-

- !i

j1 5a

5b

X3Y / \

! ij

2r10-

h 1-

6 \.

i 14 5

+ Oh6

(kg/cm2)10

Fig. 10.20 Stress paths for various loadings.

3. As the sand expands, the resistance to straining

decreases.4

. This decrease is most marked in the densest speci-mens.

We shall return to a discussion of these very importantfacts in Chapter 11.

4

Ultimate Condition

In the ultimate condition, the interlocking betweenthe soil particles has decreased to the point where con-tinuous shear deformation can continue without further

volume change. The void ratio at this stage is independentof the initial void ratio before shearing was commenced.

Test A'= 1

P= y (kg/cm2)

Fig. 10.21 Stress paths for nonfailure loadings.

PreshearTest Type

Compression loadingCompression loadingCompression unloadingCompression unloadingCompression unloadingCompression unloadingExtension loadingExtension loadingCompression unloadingExtension loadingCompression unloading

Vertical strain, e, (%)

Fig. 10.22 Stress-strain data for various loading conditions.

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PART IV

Soil with Water-No Flow or Steady Flow

Pore water influences the behavior of soil in two ways:by affecting the way in which soil particles join togetherto form the mineral skeleton (chemicalinteraction) and byaffecting the magnitude of the forces transmitted throughthe mineral skeleton [physical interaction). Chemicalinteraction has been discussed in Part II. Part IV

introduces the concepts needed to understand physicalinteraction.

Part IV specifically considers situations in which thepore pressures within a soil mass are determined by thepore pressures at the boundaries of the mass and areindependent of applied loads. Such situations existwhenever loads are applied slowly compared to the rateof consolidation and at some time (long compared toconsolidation time) after a rapid loading. Situations inwhich pore pressures are influenced by loadings will becovered in Part V.

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CHAPTER 16

The Effective Stress Concept

Our intuitive glimpse of soil behavior in Chapter 2alerted us to this fact: the behavior of a chunk of soil is

related to the difference between total stress and porepressure. The present chapter examines this concept,one of the most essential to soil mechanics.

16.1 STRESSES IN THE SUBSOIL

Chapter 8 presented equations for determining thevertical geostatic stresses in dry soil. The same equationscan also be used to determine the total vertical geostaticstress for wet soil. The soil unit weight contributing tothese total stresses is, of course, the total unit weight andthe equations corresponding to Eqs. 8.2 to 8.4 become,respectively,

= t (16.1)a

v = \Zytdz (16.2)= Xy

242 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

Groundwwa\v/sWASVAWA\V/a\va\v/av;ava surface

- Element A

Water' table

mmmm m

\ cNv

Stresses at A

Soil profile

Fig. 16.1 Stresses in the subsoil.

2. The effective stress controls certain aspects of soil

behavior, notably compression and strength.

This chapter is concerned with the first statement of theeffective stress principle, and other portions of this booktreat the second statement extensively.

The computation and portrayal of effective stress in asoil profile are illustrated in Example 16.1. Followingare several features of the stress computation and por-trayal that should be emphasized:

1. Employing units of meters and metric tons, and a

stress scale and elevation scale that are numericallythe same, the plot of static water pressure withelevation is a straight line having a slope of 45 withthe horizontal.

2. p = p + u.

3. q can be computed from either total stresses or

effective stresses:

g ~ g*=

+ ") - (ah + u)=

dv - 5hq 2 2 2

4. The coefficient of lateral stress (Eqs. 8.5 and 16.6)

refers to effective stresses not total stresses. (Thestress paths shown in Fig. 8.11c are, in fact, foreffective stresses. For the special case of zero porepressure-the situation of Fig. 8.11-the stresspaths are also for total stresses.)

16.3 PHYSICAL INTERPRETATION OF THEEFFECTIVE STRESS EQUATION

Section 16.2 defined the effective stress (Eq. 16.5) asthe total stress minus the pore pressure. While thisstatement is all that is needed to solve most soil engineer-ing problems, a physical interpretation of the effectivestress equation helps in understanding soil behavior.This section presents a physical development of theeffective stress equation.

Figure 16.2 considers on submicroscopic scale ahorizontal surface through soil at any given depth. Atruly horizontal surface cuts through many mineralparticles as suggested by Fig. 16.26, which is similar tothe view shown in Fig. 8.2. As has been discussed atlength in the preceding parts of this book, stress condi-tions at the contacts of particles, rather than withinparticles, are of primary concern to a consideration ofstrains and shear resistance within a soil mass. We are

therefore really interested in a "horizontal" surfacewhich goes through points of contact. Such a surface istermed a "wavy surface." As indicated in Fig. 16.2c,the mineral-mineral contact on the wavy surface is asmall fraction of the total surface area.

Figure 5.1 showed an average interparticle conditionfor a soil mass and thus represents the whole of the wavysurface. An element d . d corresponds to a unit cell in acrystal; it is, in effect, the repeating unit. The conditionis general as it includes all the possible interfaces,namely: mineral-mineral, air-mineral, water-mineral.

water-water, air-air, and air-water. The forces actingbetween the particles are shown in Fig. 5.1 and defined inChapter 5. The forces acting across the surface must beequal to the stress multiplied by the area:

add = Fm + Fa + Fw + R' -A'

01

odd = aAm + uaAa + uwAw + R' - A'

where

Fm

= (jAm, Fa = uaAa, Fw = uwAw

or

a = aam + uaaa + iiwaw + R - A (16.7)

where \

d-d'

" d d ' ~ d d 'R' >A'

Equation 16.7 is a statically correct relationship of thestresses acting normal to any given plane. The limita-tions to the use of this equation lie in the evaluation ofthe terms. Let us examine the terms.

In granular soils the contact stress a is usually veryhigh and the contact area ratio very small. Figure 5.16suggested a c value of 49,000 psi for a equal to 14.7 psi.The contact stresses in heavily loaded granular soils canexceed the particle crushing strength-which may be as

Example 16.1

Ch. 16 The Effective Stress Concept 243high as 1,150,000 psi. Experimental work suggests valuesof a

m for granular soils generally less than 0.03 andprobably less than 0.01 (sec Bishop and Eldin, 1950).On the other hand, in montmorillonite o and n

iucan

both be zero (see Chapter 5).In saturated soil the terms ii

aand

244 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

Horizontal surface

Wavy surface.

through contacts

Vertical view ofminerals cut byhorizontal surface.

(b)

0Vertical view ofmineral contactarea encounteredby wavy surface.

(c)

Fig. 16.2 Contact surface.

the electrical repulsive pressure R plus the pore waterpressure is the total fluid pressure existing at the mid-plane between the adjacent particles. The nature of theelectrical forces between soil particles has been discussedin preceding chapters.

As has been pointed out, Eq. 16.7 is a general andcorrect expression of equilibrium of stresses actingnormal to a given plane. For certain soil systems, or forsystems where we can justifiably make approximations,there are special cases of Eq. 16.7. Some of these specialcases follow.

1.

Saturated soil:

a = aam + uwaw + R - A (16.8)

2.

Saturated soil with no mineral-mineral contact:

a = uw + R - A3

.

Saturated soil with no net R - A:

a = aam + uwaw

or

a = oam + uw(\ - am)

(16.9)

(16.10)

(16.10a)Equation 16.8 holds for all saturated soil; Eq. 16.9 holdsfor highly plastic, dispersed systems such as mont-morillonite (Fig. 5.166); Eq. 16.10a holds for granularsoils.

On the basis of the preceding discussion, we can see atleast two circumstances where we can physically visualizethe effective stress in saturated soil:

1.

For the condition of no mineral-mineral contact:

d = R-A (16.11)2

.

For the condition of no net R - A:

a = (a - uw)avand since a is so large,

(16.12)

(16.12a)In other words, in a highly plastic, saturated, dispersedclay the effective stress is the net electrical stress trans-mitted between particles; and in a granular soil at a highdegree of saturation, the effective stress is approximatelyequal to the contact stress multiplied by contact arearatio.

The foregoing discussion helps us to understand thateffective stress is closely related to the stress transmittedthrough the mineral skeleton. For this reason

,

a is often

called intergranular stress. These detailed physicalarguments lead to some disagreement as to the exactrelationship between the stress in the mineral skeletonand effective stress (e.g., see Skempton, 1961). Be this asit may, d as defined by Eq. 16.5 has proved to be the keyto the interpretation of soil behavior.

Let us now return to Eq. 16.7 and consider partiallysaturated soil. If the air in a partially saturated soilexists within bubbles the preceding discussion onsaturated soil applies since we can take our wavy surfacearound the bubbles thereby avoiding air on our contactsurface. If, however, there is an air channel

,

as shown

2d . 2i

where

ui ~ u2 = osmotic pressure = RR = ReTi - /j2)

Rg = gas constantT = absolute temperature

= concentration of ions at 1nz = concentration of ions at 2

Fig. 16.3 Pore pressure.

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Ch. 16 The Effective Stress Concept 245in Fig. 5.1, the air must be included in the equation offorces.

Let us define the effective stress in partially saturatedsoil as

c = g - u* (16.13)where- u* is an equivalent pore pressure and

k* = uaaa + uwaw (16.14)Since

am 0, aa + aw & 1

and

u* ua + aw(uv, - ua) (16.15)and-' \

a = a - ua + aw(ua - uw) (16.16)Equation 16.16 is the form of effective stress equation forpartially saturated soil proposed by Bishop et al (1961).

In a general way, the principle of effective stress clearlyapplies to saturated soils. More research will be neededto learn whether Eq. 16.16 is really a useful quantity forthe detailed interpretation of the behavior of partiallysaturated soils.

16.4 CAPILLARITY IN SOIL

There is much evidence that a liquid surface resiststensile forces because of the attraction between adjacentmolecules in the surface. This attraction is measured bysurface tension, a constant property of any pure liquidin contact with another given liquid or with a gas at agiven temperature. An example of this evidence is thefact that water will rise and remain above the line of

atmospheric pressure in a very fine bore, or capillary,tube-. This phenomenon is commonly referred to ascapillarity.

Capillarity enables a dry soil to draw water to eleva-tions above the phreatic line; is also enables a drainingsoil mass to retain water above the phreatic line. Theheight of water column a soil can thus support is calledcapillary head and is inversely proportional to the size ofsoil void at the air-water interface.2 Since any soil has analmost infinite number of void sizes, it can have an

almost infinite number of capillary heads. In otherwords, the height of water column which can be supportedis dependent on the size of void that is effective. There istherefore no such thing as the capillary head of a soil;there are limiting values of capillary head which can bestbe explained by the setup in Fig. 16.4.2 The height of rise hc in a capillary tube is

where y = unit weight of the liquidTs = surface tension of liquida = contact angle made between the liquid and the tubeR = radius of the tube

in ea Line AISI

0)

a>'

in

LineB

Phreatic ine

1000Degree of saturation (%)

b)(a)

Fig. 16.4 Capillary heads in soil.

Figure 16.4fl shows a tube of cohesionless soil; part(b) shows a plot of degree of saturation against distanceabove the phreatic line. If the tube of soil were initiallysaturated and allowed to drain until a static conditionwere reached, the distribution of moisture could be

represented by line A. If, on the other hand, the tube ofdry soil were placed in the container of water, line Bwould represent the distribution of moisture equilib-rium.3 The lines A and B represent the two limitingconditions of capillary moisture distribution for the tubeof soil shown.

It would seem logical that point a on the drainagecurve (Fig. 16.4) is the highest elevation to which anycontinuous channel of water exists above the free watersurface. This distance therefore is taken as the maximum

capillary head hcx. Another critical point on the degreeof saturation curve for a draining soil is the highestelevation at which complete saturation exists (point b,Fig. 16.4). The distance from the free water surface tothis point is called the saturation capillary head hcs.

On the distribution curve for capillary rise there arealso two critical points. The distance from the free watersurface to the highest elevation to which capillary waterwould rise (point c. Fig. 16.4) is called the capillary riseh

cr.The distance from the free water surface to the

highest elevation at which the maximum degree of

3 The times required for equilibrium to be reached depend greatlyon the soil. Extremely long times are generally required to obtainline .

246 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

saturation exists (point d, Fig. 16.4) is named theminimum capillary head hcn.

The four capillary heads just described are limits ofthe possible range of capillary heads a soil may have.Any capillary head associated with drainage would liebetween hcx and /; , and any associated with capillaryrise would lie between hCT and hcn. Since the size of voidat the air-water interface determines the capillary head,it is reasonable, in the case of a dropping water column,for a small void to develop a meniscus which can supportthe water in larger voids below its surface, although itcould not raise the water past these larger voids. Thath

cx should be greater than hCT, and hcs be greater thanh

cn, therefore, is to be expected.Between the two extremes Ji

cx and hcn there exist manycapillary heads. The effective capillary head for any soilproblem involving capillarity would depend on theparticular problem, but would lie within the range oflimiting heads described above. For comparing varioussoils and for certain drainage problems, the saturationcapillary head /; is of much value. Also, hcs is easilymeasured.

Table 16.1, presenting test data obtained by Lane andWashburn (1946)

, indicates the range of capillary headsfor cohesionless soils.

Table 16.1 Capillary Heads

Soil

ParticleSize

D10 (mm)VoidRatio

Capillary Head(cm)

her 1}cs

Coarse 0.

82 0.

27.

5.

4 6.

0Gravel

Sandy 0.20 0.45 28.4 20.0Gravel

Fine 0.

30 0.

29 19.5 20.0Gravel

Silty 0.06 0.45 106.0 68.0Gravel

Coarse 0.

11 0.

27 82.0 60.0Sand

Medium 0.

02 0.

48-0.66 239.6 120.0Sand

Fine Sand 0.

03 0.

36 165.5 112.0

Silt 0.

006 0.

95-0.93 359.2 180,

0

From Lane and Washburn, 1946.

The fact that water held by capillarity exists at anabsolute pressure less than atmospheric pressure

,i.e.,

negative gage pressure, is an essential concept. Toillustrate this concept, let us consider the pressure in the

f

Fig. 16.5 Pressure in capillary water.,

pore water-pore pressure-at two points in the soiicolumn shown in Fig. 16.5. At these two points, point 1and point 2, are installed standpipes, as shown. Thtwater in the open ends of both standpipes is at the leve'of the phreatic line; therefore the pore pressure at poin11 is negative and numerically equal to the verticadistance from the phreatic line to point 1 multiplied b}the unit weight of water, and the pore pressure at pointis also minus and equal to the vertical distance fronpoint 2 to the phreatic line multiplied by the unit weighof water.

Whether the column of soil shown in Fig. 16.5 wa'initially dry and imbibed water or was initially saturateand drained has no influence on the pore pressure at ;given point. The pore water pressure at any elevation iithe soil column is equal to the height of that point abovithe phreatic line multiplied by the unit weight of wateas long as there is continuous static water.

Figure 16.4 shows that there is no unique relationshi]between degree of saturation and height above thphreatic line; the relationship depends on the history cthe sample. Figure 16.5 shows that for static equilibriur(and continuous channels of water) the pore watcpressure at any point i?,exactly equal to the height cthe point above the phreatic line multiplied by the uniweight of water, regardless of the degree of saturatioi

Combining these two facts, we can conclude that porewater pressure is not a unique function of degree ofsaturation but depends also on the history of thesample.

16.5 THE COMPUTATION OF EFFECTIVESTRESS FOR STATIC GROUND WATERCONDITIONS

Example 16.1 illustrated the computation of stresswithin a highly idealized subsoil. Example 16.2illustrates the computation for an actual soil profile at anindustrial site in Kawasaki, Japan. Underlying 15 m ofsand and silt are three strata of recently depositedsedimentary clays. The three clays are composed of thesame minerals but were deposited under slightly differentgeological conditions and thus have different engineeringproperties. Underlying the three clay strata is a stratumof dense, clean sand.

A wall has been constructed around the property and acanal has been dredged in-order that tankers may dockat a pier which runs perpendicular to the wall. The sitedevelopment plan calls for a number of product storagetanks to be constructed near the canal. The sketch showsone of these tanks; it is 30 m in diameter, 13.8m in

height, and has a nominal storage capacity of 10,000 kl.This planned construction presents both a stabilityproblem and a settlement problem which must beinvestigated. The factor of safety against a shear rupturethrough the soil beneath the tank must be adequate;the settlement of the tank must be predicted in order thatproper allowance be made for piping connections to thetank. As we shall see later in this book, an early andessential step in both a stability and a settlement analysisis the determination of the effective stress existing in thesubsoil prior to planned construction.

Shown in the sketch are plots of vertical total stressav, pore pressure u, and vertical effective stress dvagainst elevation. Also shown is a table of computationsfor the stresses. The stresses between elevations +4.0and +3.3 are based on full saturation and full capillarity.

The pore water pressure plotted is for the conditionsof no ground water movement and the phreatic surfaceat elevation +3.3. The phreatic surface was taken as thelevel to which the water rose into holes excavated into

the sand-silt or into the boreholes.4 The assumption ofno ground water movement is very common althoughnot always sound, especially for a site such as Kawasakiwhere there has been recent placement of fill. Further,in the Kawasaki area wells have been installed into the

deep sand stratum in order to secure water for industrialuses. Because of the recent site filling and because ofpumping from wells in the area, the pore pressure plot4 One must wait until the water level becomes stationary for thistechnique to give the correct phreatic surface.

Ch. 16 The Effective Stress Concept 247and dependent effective stress plot are suspect. Porewater pressure measuring devices (PI to Pb) weretherefore installed at the elevations shown.

If the devices had shown that the pore pressure wasindeed equal to the distance from the device to thephreatic line multiplied by the unit weight of water, theplots of pore pressure and effective stress shown inExample 16.2 would be correct. In fact, however, thereis considerable deviation in pore pressure from the staticpressure plot and the assumption of a static condition isvery poor for the Kawasaki site. The actual porepressure measurements are presented in Chapter 17, andthe details of computing the correct pore pressures andeffective stresses are discussed there. Example 16.2illustrates how the pore pressure plot for a static case isdetermined.

The increment of total stress Aav was obtained bymultiplying the thickness of the stratum by the total unitweight of the soil. Thus the total weight used for thiscomputation is the average for the stratum. The factthat only one void ratio or unit weight value is given foreach stratum should not lead one to infer that thesevalues are constant for the entire stratum. Fig. El6.2-2presents the results of a compression test on a sample5 ofClay II. This plot, in the form of effective stress 5V on alog scale against void ratio e, is a straight line. As notedon the sketch, the total unit weight of Clay II actuallyvaries from 1.49 tons/m3 at the top of the clay stratum to1.51 at the bottom, a very small variation. The use of

1.50 as a total unit weight for the entire stratum is thus

acceptable. If, however, the actual variation in effectivestress of 7.5 tons/m2 had occurred in going from 2.0 to9.

5 rather than 19.6 to 27.2, the variation in void ratiowould have been 0.67 rather than 0.15. In other words,we must remember that generally variations in void ratioare much larger for a given stress increment at a lowstress level than for a high stress level. Where there is amarked decrease in void ratio with depth in a particularstratum, the plot of total stress with depth should not beplotted as a straight line, but rather as a curved one witha larger increment of total stress for a given change inelevation as one proceeds deeper. In such a situation,the clay stratum can be divided into several layers withthe contribution of each layer to total stress determined.

Example 16.3 illustrates the determination of stressesin a nonsaturated soil. A column of fine, uniform sand

at constant void ratio will be used. The soil was initiallysaturated and then permitted to drain with the phreaticline maintained at elevation zero. After static equilibriumhad been reached, measurements of degree of saturationwere made at a number of points in the sand and a plot ofpercent saturation versus elevation was prepared. Tothe right in the example are plotted various stresses as afunction of elevation.

5 The e versus log dv plot has been extended back to small slresses.

Example 16.2

+ 10r-

E -20

-40

10,000 Kltank

+ 2, High tideC

c= 0

.

254 I \ \ / 0, Low tideyt = 1.80 t/m3c= 1.10 d, = 0.6 cm /secA = 3 x 10~3cm/sec

Sand-Silt

7/ = 1.70e = 1.50

Cc = 0.600,,= 2 x lO"3

Clay Ik = 2 x 10

Tt = 1.501 e = 2.20

k = 2 x lO"7

Cc = 1.00= 2 x lO-3

Clay II

7t= 1.75.

6=1.10 a = 0.45A = 4x 10"7 cv = 4 x lO-3

Clay III

Canal

40 60Stress (ton/m2)

Elevation Soil yt Az Act, u 0V

+4.0 0 -0.

7'

0.

7

Sand-Silt i.pp 0.7 1

.

3

+ 3.3 1.

3 0 1.

3

Sand-Silt 1.

80 14.3 25.7

-11 27.0 14.3 12.7

Clay I 1.70 10 17.0

-21 44.0 24.3 19.7

Clay II 1.50 15 22.5

-36 66.5 39.3 27.2

Clay III 1.75 15 26.2

-51 92.7 54.3 38.4

Fig. El6.2-1 Kawasaki subsoil profile.

248

Example 16.2 continued

3.

2

3.

0

2.

8

.5 2-6>

2.

4

2.

2

2.

0

ay II-

Comfsession curve-cG = 2.60

2.28, 7t = 1-49

2.20, 7r= 1-50

2.

13iKi ii i!

,

1

n -

4 5 6 7 8 9 10 20 30 40 50 60Effective stress, au (ton/m )

Fig. E16.2-2

Example 16.3

200

o

II

150

100

50

20 40 60 80Percent saturation

200

150

?

.

1 ioo

50

0

i

-

._

- ji

\ j //-

- pi

---

-_ _

_

-

\ I

/Vi

p- ..

\ \ -- --1/..

_

tftlK\

\s__

-

-

.

__

NO \

NA-

-

AS./ \ N

.

.

_

1

\..

..

_

\-

_

__

100 0 100

Fig. E16.3 Stresses in a nonsaturated soil

200

Stress (g/cm2)300 400

249

250 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

The plot of elevation versus pore water pressure is astraight line going from zero at the phreatic line, elevation0

, to minus 180g/cm2 at elevation 180 cm. The totalvertical stress ov is the total weight of soil and water perunit area above the point in question. Since the degreeof saturation varies with elevation, the total unit weightof soil varies with elevation, and the plot of total stresswith elevation is not a straight line.

At the far right is plotted the difference between totalstress and pore water pressure. It was obtained merelyby subtracting the pore water pressure from the totalstress; since the pore water pressure at all points isnegative, the numerical value of the pore pressure wasadded to that of the total stress. Also shown is the plotof a - (Sl\00)uw, i.e., the effective stress as defined byEq. 16.16, with ua = 0 and aw = S/IOO.

At some height in a column of fine sands, such as thatin Example 16.3, the pore water ceases to be continuous.When the pore water is not continuous, the pore waterpressure is no longer a unique function of height abovethe phreatic line. Water could be trapped in voids farabove the phreatic line and could still exist at positivepressures.

As can be seen from Example 16.3, there is considerabledifference between the values of a - uw and a -(Sl\00)uw above elevation 40. The stress a - {S/\00)uwis probably closer to the stress that best correlates withsoil behavior than is cr - iiw.

16.6 SUMMARY OF MAIN POINTS

1.

The effective stress is:a

.

Saturated soil:5 = a - u

b. Partially saturated soil:

a o - ua + aw(ua -

2. For geostatic stresses and static pore water:

a- (yv = 2 yt Ar.

b. z/ = z y .

c. dh = Kdv.

3. In granular soils, the effective stress is approxi-

mately the contact stress multiplied by the ratio ofcontact area to total area.

4. In general, the effective stress is essentially theforce carried by the soil skeleton' divided by thetotal sectional area.

5. Within the capillary zone of continuous moisturethe pressure is negative, and at any point it isnumerically equal the height of the point above thephreatic line multiplied by the unit weight of water.

PROBLEMS

16.1 For the Thames Estuary clay in Fig. 7.10 make aplot of a,,, u, and 5V versus elevation for a depth of 40 ft.(Compare your plot of 6V with that given in Fig. 7.10.)

16.2 Refer to the subsoil profile for the South Bank inFig. 7.8. Which would be the larger value of dv at a depth of140 ft:

a. 6V computed for static pore pressure.

b. dv computed for the pore pressure indicated by the

standpipe in Fig. 7.8.What is the magnitude of the difference between the two

values of dvl

16.3 The water table in Example 16.1 rises 2 m (the tidecomes in) while the soil surface elevation remains constant.Compute the values of av, dv, u, ah, 6h at element A.

16.4 The contact area ratio in the sand under the center of

the tank at elevation -5 m in Example 16.2 is 0.1 %. Discussthe likelihood of sand particle crushing following the fillingof the tank with water.

16.5 The height of soil column in Fig. 16.4a is 2.0 m.The soil (fine sand) has the following properties:

e = 0.473 (constant with depth)G = 2.69

aw 100%=0

.05%.

For the fully drained condition (line A) compute forpoint a:

Uw, "*t 0v, dv> 5V

Hint. The air pressure is atmospheric. Assume geostaticstresses. *

16.6 Refer to Fig. 5.166. Compute the effective stress foran interparticle spacing of 40 A. The spacing is held constantwhile salt is added to water around the particles. Does theeffective stress increase or decrease? Explain.

CHAPTER 17

One-Dimensional Fluid FlowV

17.1 THE NATURE OF FLUID FLOW INSOILS

Chapters 17, 18, and 19 deal with the flow of fluids-primarily water-through soil. The engineer mustunderstand the principles of fluid flow in order to solveproblems {a) involving the rate at which water flowsthrough soil (e.g., the determination of rate of leakagethrough an earth dam); (b) involving compression (e.g.,the determination of the rate of settlement of a founda-tion); and (c) involving strength (e.g., the evaluation offactors of safety of an embankment). The emphasis inthese three chapters is on the influence of the fluid on thesoil through which it is flowing; in particular, on theeffective stress.

In general, all voids in soils are connected to neighbor-ing voids. Voids that are isolated from neighboringvoids are impossible in an assemblage of spheres, regard-less of the type of packing. In the coarse soils-gravels,sands, and even silts-it is hard to imagine isolatedvoids. In the clays, consisting as they usually do ofplate-shaped particles, a small percentage of isolatedvoids would seem possible. Electron photomicrographsof natural clays, however, suggest that even in the finestgrained soils, all of the voids are interconnected.

Since the pores in soil are apparently interconnected,water can flow through the densest of natural soils.Thus in a tuoe of soil

,such as that illustrated in Fig. 17.1,

water can flow from point A to point B. Actually; thewater does not flow from point A to point B in a straightline at a constant velocity, but rather in a winding pathfrom pore to pore, as illustrated by the heavy line in Fig.17.1. The velocity of a drop of water at any point alongthe winding path depends'on the size of the pore and itsposition in the pore, especially on its distance from thesurface of the nearest soil particle. However, in soilengineering problems, the water can be considered to

flow from point A to point B along a straight line at aneffective velocity.

17.2 DARCY'S LAW

In the 1850s, H. Darcy, working in Paris, performed aclassical experiment. He used a setup similar to thatshown in Fig. 17.2 to study the flow properties of waterthrough a sand filter bed. He varied the length of sampleL and the water pressure at the top and bottom of thesample; he measured the rate of flow Q that passedthrough the sand. Darcy experimentally found that Qwas proportional to {hz - /iJ/L, and that

Q = k] - A = kiA (17.1)where

Q = the rate of flowk - a constant, now known as Darcy's coefficient of

permeability1h2 = the height above datum which the water rose in a

iUandpipe inserted at the entrance end of thefilter bed

h4 = the height above datum which the water rose in astandpipe inserted at the exit end of the filter bed

L = the length of sampleA - the total inside cross-sectional area of the

sample container

i = - the gradient

Equation 17.1, now known as Darcy's law, is one ofthe cornerstones of soil mechanics. During the centurysince Darcy performed his monumental work, Eq. 17.1has been subjected to numerous examinations by manyexperimenters. These tests have shown that Darcy's lawis valid for most types of fluid flow in soils. For liquid

1 Chapter 19 treats this important soil property, permeability.Values given in Chapter 19 show a very wide range, from greaterthan 1 cm/sec for gravel to less than 1 x 10-7 cm/sec for clay.

251

252 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

Flow path-macroscopic scaleFlow path-microscopic scale

Fig. 17.1 Flow path in soil.

flow at very high velocity and for gas flow at very low orat very high velocity, Darcy's law becomes invalid. Thevalidity of Darcy's law is treated later in this chapter.

17.3 FLOW VELOCITY

A further consideration of the velocity at which a dropof water moves as it flows through soil helps to under-stand fluid flow. Equation 17.1 can be rewritten as

= = , (17.2)

Since A is the total open area of the tube above the soil(Fig. 17.2), v is the velocity of downward movement of adrop of water from position 1 to position 2. This velocityis numerically equal to ki; therefore k can be interpretedas the approach velocity or superficial velocity for agradient of unity, i.e., k = vji or k - v for a gradientequal to 1.

Sand

Qout_

2_

'

"--Datum

Fig. 17.2 Darcy's experiment.

From position 3 to position 4 in the sample shown inFig. 17.2, a drop of water flows at a faster rate than itdoes from position 1 to 2 because the average area ofchannel available for flow is smaller. This reduced areaof flow channel is represented in Fig. 17.3, which is thetest setup in Fig. 17.2 with the mineral and void volumesseparated. Using the principle of continuity, we canrelate the velocity of approach v to the average effectivevelocity of flow through the soil vs as follows:

Q = vA = vsAv

VA ALv

.

- v - = v = VA

.yAJL V.v

n

The average velocity of flow through the soil vs, termedthe seepage velocity, is therefore equal to the velocity ofapproach divided by the porosity:

t; - - - (17.3)

Equation 17.3 gives the average velocity of a drop ofwater as it moves from position 3 to position 4; this isthe straight line distance from 3 to 4 divided by the timerequired for the drop to flow from 3 to 4. As notedearlier in this section, a drop of water flowing throughthe soil actually follows a winding path with varyingvelocity; therefore i's is a fictitious velocity for an assumeddrop of water that moves in a straight line at a constantvelocity from position 3 to position 4.

Even though approach velocity and seepage velocityare fictitious quantities, thefy both can be used to computethe time required for water to move through a givendistance in soil, such as between positions 3 and 4.

17.4 HEADS

In the study of fluid flow it is convenient to expressenergy, both potential and kinetic in terms of head, which

Fig. 17.3 Flow channel.

Ch. 17 One-Dimensional Fluid Flow 253

is energy per unit of mass.2 The following three headsmust be considered in problems involving fluid flow insoil:

1. Pressure head, hv = the pressure divided by the

unit weight of fluid.2

. Elevation head, he = the distance from the datum.3

. Total head, h = hv + he - the sum of pressurehead and elevation head.

When dealing with flow through pipes and openchannels, we must also consider velocity head. Thevelocity head in soils, however, is much too small to beof any consequence and thusv can be neglected. (Forexample, a high velocity for fluid flow in soil is 2 ft/min,a value which gives a velocity head of 0.00002 ft. Thishead is far less than the accuracy with which the engineercan normally measure pressure head or elevation head.)Those engineers dealing with pipe and channel flow definetotal head as velocity head + pressure head + elevationhead; and they define piezometric head as pressurehead

.

- elevation head. For flow through soil with itsnegligible velocity head, the total head and piezometrichead are equal.

Since both pressure and elevation heads can contributeto the movement of fluid through soils, it is the totalhead that determines flow, and the gradient to be used inDarcy's law is computed from the difference in totalhead. The importance and truth of this statement can beseen from the two situations shown in Fig. 17.4.

Figure 17.4tf shows a common bucket full of water in astatic condition. In Fig. 17.4 the heads for the twopoints, number 1 near the top and number 2 near thebottom of the bucket, are labeled and plotted. Betweenpoints 1 and 2 there exists both a pressure gradient andan elevation gradient; there is, however, no gradient oftotal head since the total head of the two points isidentical and equal to h. Even though the pressure atpoint 2 is considerably greater than that at point 1,water does not flow from point 2 to point 1; further,

2 Kinetic energy =Mv2 ML2r22 r2L

= ML

Elevation energy = MLpM MM L3

Pressure energy = = -jy~ ~ = MLHead =

P

energy

M

ML

M= L

where

M = mass

v = velocityg = acceleration of gravityL = lengthT = time

p = pressure'p = density

flow does not occur from point 1 toward point 2 eventhough the elevation head at point 1 is three times that atpoint 2.

Figure 17.46 shows a capillary tube in which there iswater standing to a height of hc. As with the water in thebucket there is no total head gradient.

These very simple examples illustrate two importantprinciples:

1. Flow between any two points depends only on the

difference in total head.2

. Any elevation can be selected for datum as a baseof elevation heads. The absolute magnitude ofelevation head has little meaning; it is the differencein elevation head that is of interest and the differenceof elevation head between any two points is thesame regardless of where datum is taken.

17.5 PIEZOMETERS

In soil mechanics we are especially interested inpressure head, since the pore pressure needed to computeeffective stress can be obtained from the pressure head.The pressure head at a point can be either measureddirectly or computed using principles of fluid mechanics.

Pore Pressure Measurements in the Laboratory

The pressure head or water pressure at a point in asoil mass is determined by a "piezometer," a wordliterally meaning "pressure meter." Figure 17.5 showstwo simple piezometers. On the right side of the soilsystem is a manometer or standpipe connected to aporous tip located at midheight in the soil system. Onthe left of the tube is a regular Bourdon gauge connectedto the porous tip.

As illustrated in Fig. 17.5, as the water flows in the soil'

sample a pressure head at its midheight of i ft is indicatedby the manometer and a pressure of 31.2 lb/ft2 by thegauge. While the two piezometers shown in Fig. 17.5 aresimple in principle, they both have a serious drawback:a flow of water from the soil into the measuring system isrequired to actuate each device. The drawback can beillustrated by considering what would happen if thereservoir height were increased 1 ft. This would increasethe pressure head at midheight by I ft> meaning that avolume of water equal to i ft times the inside area of themanometer tube would flow from the soil into the man-

ometer. Even though considerably less flow would berequired to actuate the gauge, a measurable amount ofwater would flow from the soil into the gauge. If the soilin the permeameter were pervious, the time required forre-establishment of an equilibrium flow condition wouldbe minor. If, however, the soil were a silt or clay, anappreciable amount of time would be required for thewater to flow from the soil into the manometer or

254 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

7\\1\

I

\Elevation = 0 /Datum Head

Elevation PressurePoint Head Head Total Head

1 hpi Kx + hvx = h

2 he* hM he2 + }h2 = h

1 \z /V. \ /9.

7,

toCD

V,

7,7,7,7,7,7s7, / -A7,7,7,7.

7, \5r7,7, /g 2

DatumHead

Elevation PressurePoint Head Head Total Head

1 K -K /;c - hc = 0

2 0 0 0

(6)Fig. 17.4 Heads in static water, (a) In container. (6) In capillary tube.

pressure gauge. The two systems shown in Fig. 17.5would therefore involve large time lags in measuringpore pressures in relatively impermeable soils. Tomeasure pore pressures under "no-flow" conditionsvarious types of piezometers have been developed (see,for example, Lambe, 1948; Bishop, 1961; Whitman etal.1961; and Penman, 1961).

Figure 17.6 shows a most interesting and'importanttest result. A stress of 20 psi was applied to a saturatedclay specimen and held until the pore water pressure was

zero. The stress was instantaneously removed and theresulting negative pore pressure was recorded. Therecord shown in Fig. 17.6 indicates an almost immediatepore pressure response equal to the magnitude of therelieved stress, 20 psi. This measured pore pressure was5

.3 psi below absolute zero. The pore water sustainedthis pure tension for a short period of time, but cavita-tion (probably in the measuring system) resulted in arecorded pressure of -14.7 psi; i.e., absolute zero. Fromother measurements we can infer that pore pressures

Ch. 17 One-Dimensional Fluid Flow 255

in

II11

o-44)Bourdon

Open manometergageor standpipe

1iPressure head, (ft)

Fig. 17.5 Flow-type piezometer.

below absolute zero can exist over long periods of time,and that these negative pore pressures may be as largeor larger than 60 psi.

The measurement of pore water pressure, especiallyvalues below atmospheric pressure, in partially saturatedsoil requires special precautions. From the nature of themeasuring systems described, it is apparent that airmust be kept out of the measuring devices. The easiestway to avoid difficulty with entrapped air is to use asensing element which tends to prevent air from entering.If, for example, the sensing element is a fine porous stoneinitially saturated with water, capillarity will keep airfrom entering the stone until the air pressure exceeds thecapillary pressure, or breakthrough pressure, of thestone. Porous stones with breakthrough pressures ashigh as 60 psi are commercially available. Such highbreakthrough pressure stones are desirable for measuringpore water pressures in partially saturated soils.

Acr = -20 lb/in2

7/

1

0 1 C 2 3 4 5Time (sec)

Fig. 17.6 Pore pressures measured by transducer piezometer.

Pore Pressure Measurements in the Field

The preceding subsection discussed the measurementsof pore water pressure during laboratory tests. Suchmeasurements are usually helpful (and often essential) ininterpreting the test data. The basic principles involvedin laboratory measuring systems can be used in fieldpiezometers. Figure 17.7 shows a modified Casagrandepiezometer. This piezometer was developed by A.Casagrande. It consists essentially of a porous ceramicstone inserted in a plug of clean sand. Since water must

flow from the sand into the plug and up the tube toregister an increase in pore water pressure, it is a "flow"piezometer. By having a relatively large zone (the sandsurrounding the porous point) from which to drawwater, this piezometer requires relatively short time lags.The Casagrande piezometer has been widely and success-fully used in civil engineering field installations.

A crucial feature of any field piezometer is the necessityof sealing the sensing unit into the zone where porepressure is desired.

Since flow of water from the soil at the point wherepore pressures are being measured is required to actuatethe Casagrande piezometer, it would not be a satisfactorydevice to measure rapidly changing pore water pressures.We can and should determine the response time of aCasagrande piezometer by filling the tube with water anddetermining the time required for attaining a steadyreading (see Hvorslev, 1949).

The reader is referred to the Proceedings of theConference on Pore Pressure and Suction in Soils (1961)for more information on piezometers.

17.6 CALCULATION OF PRESSURE HEAD

The principles of flow through porous media, whichhave been described in the preceding sections, can bemade clearer by considering Figs. 17.8 to 17.11.

Figure 17.8 shows a tube of soil having a porosity of0.33 and a permeability of 1 ft/min in which water is

flowing vertically downward. Atmospheric pressure ismaintained at the top of the reservoir water (elevation 12)and at the bottom of the tail water (elevation 0). Datumis taken at the tail water; this selection is merely one ofconvenience, since any location can be selected fordatum. To the right of the tube are plots of heads andvelocity versus elevation.

In general, is it more convenient to first determine theelevation and total heads and then compute the desiredpressure head by subtracting the elevation head from thetotal head. The elevation head is merely the elevation ofthe water at any point under consideration. Since thehorizontal scale for head has been taken the same as the

vertical scale for elevation, the slope of the elevationhead versus elevation plot is one to one, i.e., 45. The

256 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

-Removable cover

Surface of soil

. Sand lightly tamped

4 ft Main seal

1 in. Initial seal

Plastic tube: {in. O.D.

2f Standard pipe

Rubber bushing

Piezometer tipPorous tube: ceramic material

(Norton P2120) l in. O.D., lin.l.D.

Rubber stopper

Sand

Fig. 17.7 Casagrande piezometer (not to scale).

total head at elevation 12 is the elevation head since thepressure head is 0. In flowing from elevation 12 toelevation 8 "no" total head is lost so that the total head

at elevation 8 is also 12 ft. Similarly, we note that thetotal head at both elevation 0 and elevation 2 is 0. Sincethe soil has uniform permeability and porosity thedissipation of total head in flowing through the soilmust be uniform; the total head plot is therefore astraight line running from a value of 12 at elevation 8 to avalue of 0 at elevation 2. The pressure head is obtainedby subtracting the elevation head from the total head forany point under consideration.

To draw the total head line vertically between eleva-tion 12 and 8, and between 2 and 0, assumes that the

friction loss of flow in the entrance and exit parts of thetube is negligible compared to the head lost in flowthrough the soil. The validity of this assumption iseasily checked by computing the head lost in the entranceand exit parts of the tube. Using a reasonable frictionalcoefficient and the principles of hydraulics, we computethe head lost in this 6 ft of-tube as 3 x lO"6 ft.

..

This is

indeed negligible.. The velocity of flow in the entrance and exit parts ofthe tube was computed from Darcy's law, v - k x i -1 ft/min x -ir == 2. The seepage velocity is equal to theapproach velocity/porosity, i.e., 2/0.33 = 6 ft/min.

Figure 17.9 presents no new principle except that thewater is flowing upward through the soil. Figure 17.10

Ch. 17 One-Dimensional Fluid Flow 257

12 \\

10\\

8

S 6CO LU Za*n5=

II IIZ

COz

\I2

12108VelocityHead (ft)(ft/min)

Fig. 17.8 Example of downward flow.

involves horizontal flow in which the elevation head is

constant. In Fig. 17.11 the area of the permeameter,andthe properties of the soil change at elevation 4. Since nowater is added to or removed from the system, the rateof flow of water in soil I must be equal to that in soil II.Thus

From this equation we find that the total head lost insoil I is half of that lost in soil II

,

or that a total head of4 ft is lost in soil I and 8 ft in soil II. The total head line

in Fig. 17.11 can be drawn on this basis, and the pressurehead can be obtained by subtracting the elevation headfrom the total head.

The four preceding examples of flow through porousmedia illustrate the following five interesting points.

I. The velocity head in soils is negligible. Th e m a x i m u m

velocity in the four examples was 12 ft/min. Thevelocity head for this maximum case is 6 x 10-4 ft

,

a value that is indeed negligible.2

. All head is lost in soil. Our computation for Fig.17.8 clearly indicates that in flow through tubes ofsoils we can neglect completely the head lost in theportions of tube with no soil.

3. Negative pore pressure can exist. In the examples

given in Figs. 17.8, 17.10, and 17.11 we have porepressures which are less than atmospheric, so-called

rot;o.-

"g II IIro -

CO

16

14

12

,

10

i aI

I

i6

4

2

0

\!

\\

\\

//

/-

//

/

KN \

258 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

Air pressureof 3.4 psi

11

-Sand,

n = 0.33k= 1 min/ft

11

0 2 4 6 8 10 12

6L

: 4

; 2i

: oi

!-2

-4

- 4

Length (ft)

5

\\

%

\\

Elev ation he;ad

%\

- 0

III

Fig. 17.10 Example of horizontal flow.

negative pore pressures. We already have noted inthe preceding section that pore pressures are possibleeven below absolute zero.

4. Direction offlow determined by total head difference.

The four examples present cases where flow wastoward increasing elevation head, toward decreasingelevation head, toward increasing pressure head,

and toward decreasing pressure head. These factsillustrate a point we have already made: it is thetotal head that determines flow and not pressurehead or elevation head alone.

5. Method of head determination. Our four examples

illustrate the general -lethod of determining heads:first determine elevation and total ,head

,

and then

14

12 7

10 K4. h /?> //

8

A = 4 ft2

Ch. 17 One-Dimensional Fluid Flow 259

compute the desired pressure head from these two.In a simple situation, such as Fig. 17.8, we canreadily determine the pressure head at any pointwithout knowing the elevation or total head; in amore complex situation such a procedure may notapply. In many practical examples, however, theengineer will have a measure of the pressure headand elevation head, and thus may compute totalhead from these two.

Let us summarize our discussion of heads by notingthat there are three of interest to the soil engineer:

1. Elevation head. Its absolute magnitude depends

on the location of datum.2

. Pressure head. The pressure head magnitude is ofconsiderable importance since it indicates the actual

pressure in the water. The pressure head is theheight to which the water rises in the piezometerabove the point under consideration.

3.

Total head. The total head is the sum of the

elevation and pressure head and is the only headth?t determines flow. The total head is used inDarcy's law to compute gradient.

17.7 EFFECTIVE STRESS IN SOILWITH FLUID FLOW

Chapter 16 treated the computation of effective stressfor static ground water conditions. Thus far this chapterhas considered the computation of heads for simplesituations of fluid flow. Let us now combine these two

3rCM

TSTr-1

CM

co mm co 4;d _

.

2

\\

//

/

\\

\

/

\/

//

v

\ I/

//

\0

0 1Supporting

screen

2 3Head (ft)

--

\\-

r\

1 ! \ Auf \|-200

Stress (lb/ft2)400

Ele-vation a Ob/

(ft) (lb/ft2) (lb/ft2) (lb/ft2) ft2)

3 0 0 0

1 x yw = 62.4

2 62.4 1 x rw = 62.4 0

2 x y, = 266.0

0 328.4 5yw =312.0 16.4

Fig. 17.12 Heads and stresses for fluid flow.

CHAPTER 21

Drained Shear Strength

Chapter 11 presented fundamental concepts ofstrength theory and strength data for dry cohesionlesssoil. As long as stresses are expressed in terms ofeffective stresses, the concepts and data presented inChapter 11 also hold for saturated cohesionless soilswherein there are no excess pore pressures. Chapter 21extends Chapter 11 to cover the strength behavior ofclays and compares this behavior of clays with that ofsands. As in Chapter 20, Chapter 21 is limited to thecondition where the pore pressure is either static or zero.Excess pore pressures generated by strains are permittedto dissipate by the free movement of pore water, i.e.,drained conditions.

Drained tests are usually called consolidated-drained,abbreviated CD, in order to have consistent terminologywith undrained tests presented in Part V. It is reallynecessary, however, to use only the word drained, sinceif the excess pore pressure is equal to zero at all times theclay must be consolidated at the start of the test. Sincedrained tests on clay are run at a slow rate in order topermit the excess pore pressures to dissipate, drainedtests are also called slow tests.

To present the principles of drained strength, use willbe made of actual test data on remolded Weald clayobtained by Henkel (1956, 1959). Weald clay is anestuarine clay of the Cretaceous period and has thefollowing properties:

Liquid limit = 43%Plastic limit =18%

Plasticity index = 25%Percent clay minus 0.002 mm =40%

Activity = 0.6Specific gravity = 2.74

The Weald clay is used for illustration both because ofthe wealth of high-quality test data available and becausethis clay has typical behavior. Later in this chapterdeviations from typical behavior are presented anddiscussed.

21.1 NORMALLY CONSOLIDATED CLAY

On identical specimens of remolded Weald clay we runthe following three tests:

1. Consolidated to 10 psi and then failed in a standardtriaxial test.

2. Consolidated to 30 psi and then failed in a standardtriaxial test.

3. Consolidated to 100 psi and then failed in a standardtriaxial test.

The results of these three tests are shown in Fig..,21.1.In the standard triaxial test the vertical stress a

vis the

major principal stress and the horizontal stress ah is theintermediate and minor principal stress. Figure 21.lapresents half the deviator stress, q, versus vertical strain.Figure 21.1 presents volumetric strain versus verticalstrain. Figure 21.1c shows the effective stress path foreach of the tests, and a line drawn through the peak valueof q for each test. This line, called the Kf-\ine, is a plot ofqf versus pf. Figure 21.1 d' presents void ratio versus/?0, pf, and qf.Strength Versus Effective Stress

Based on the data shown in Fig. 21.1c plus many otherpoints not shown, the effective stress-strength relation olremolded, normally consolidated Weald clay is found tobe a straight line through the origin. Thus there appearsto be no essential difference between the form of the

effective stress-strength relationship, as found fromdrained tests, of sand and of normally consolidated clay:neither soil exhibits a qf intercept at pf equal to zero.meaning neither soil exhibits "cohesion." The slope olthe Kf-\'me is a:

a = tan"1 &Pf

and (from geometry, as noted in Chapter 11)$ = sin-1 tan a

For the Weald clay a = 201 and $ = 22.304

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Ch. 21 Drained Shear Strength 305

I

5050

CM

CM

I150100 200503020 010

5 (lb/in/)(%)fa;

V

tio 0.8

0 0.

7

e-Test 2

Test3

It 10 0.

6Y 0 10 20 30

I0

.

450 100 150 200

9 p (lb/in.*)

Fig. 21.1 Test results on normally consolidated Weald clay.

Figure 21.2 shows stress-strain and stress path data fornormally consolidated Weald clay presented in anormalized fashion. Since the Kr\\nt is straight, thep-q plot can be accurately normalized. The stress-straindata can only be approximated by normalized curves.

Stress-Volume RelationshipFrom Fig. 21.1 we can see that the test data from the

drained tests on the normally consolidated Weald clayindicate that e0 - p

, ef - pn and ef - qf are smoothplots. Also shown in this figure are lines indicating howthe void ratio changes during the drained tests. Thestress-volume data in Fig. 21.\d are replotted in Fig. 21.3in a more convenient form

.

.Water contents are used in

place of void ratios since they are more convenient todetermine

,

and for a saturated soil, water content andvoid ratio are uniquely related. The stress data areplotted on a log scale since it has been found empiricallythat void ratio or water content versus the variousstresses form approximately straight lines which areparallel to one another for normally consolidated clay.

From the data plots in Fig. 21.\d or 21.3, we see thatthere is a three-way relationship among shear strength,effective stress at failure

,

and water content or void ratio

at failure. Jn Part V we shall see that the sameqf - pf - wf relation still applies to Weald clay whenthe clay is sheared without permitting water to escape.

Any factor that alters the void ratio, i.e the density ofthe skeleton, should alter the drained strength, whichshould also be accompanied by a change in the effectivestresses existing at failure. In a sense

, a change ineffective stress can produce a material with a differentdensity and hence a different strength. It is difficult tosay which factor-density or effective stress-controlsstrength: all three of these factors are interrelated.However, it is more useful to use effective stress as theprimary controlling variable.

Other Loading ConditionsThere is an infinite number of possible stress paths for

shearing a sample of clay. Figure 9.8 showed fourconvenient stress paths for the triaxial tests. In general,the strength parameters of normally consolidated soil areapproximately independent of the stress path to failure.As a first and a very good general approximation, we cantake qf versus pf for normally consolidated clay as aunique straight line. The value of qf> of course, dependsvery much on the stress path of a loading since pf is a

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306 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

lb 0.

5 0.

5II

lb=> CM

0010 20 30 0

.

5 1.

00 1.

5

10

0

1+10100 20 30

Fig. 21.2 Normalized test results.

function of loading stress path. Examples 21.1 to 21.3illustrate the estimation of for various loadingconditions.

Comparison of Clay and SandThe preceding discussion illustrates that the behavior

of normally consolidated clay during drained shear isessentially similar to that of sand, or more particularly,loose sand. That is, for both classes of soil the failurelaw is

ofS tan

28

Test 126

24

l-A Vest 2g

22N0\

20 5 b Test 318

1680 100 200 4008 10 20 401

qf, pf, and p0 (lb/in.2)Fig. 21.3 Test results on normally consolidated Weald clay.

Typical Values of for Normally Consolidated ClaysFigure 21.4 gives a good indication of typical values of

for soil. Although there is considerable scatter, thereis a definite trend toward decreasing $ with increasingplasticity. It is difficult to determine the effective stress-strength relationship for an extremely plastic clay suchas montmorillonite or thp Mexico City clay because ofthe extremely long time required for dissipation of excesspore pressures during the test. Hence the magnitude of for very plastic soils is uncertain.

21.2 OVERCONSOLIDATED CLAY

Now let us take several specimens of remolded Wealdclay and consolidate them to 120 lb/in.2 This stress(represented by pm for Vnaximum past p) is greater thanany effective stress to which the specimens have beensubjected since remolding. Now the chamber pressureacting on each specimen is reduced to values p0 rangingfrom 5 to 70 lb/in.2, and water is permitted to flow intothe specimens so that they come to equilibrium underthese reduced effective stresses. The overconsolidationratio OCR (defined as OCR = pjpo; in one-dimensionalcompression, OCR is dvmlavQ) is a convenient way tocharacterize the condition of these specimens. Thevalues of OCR for our specimens range from 1.7 to 24.Strength Versus Effective Stress

Figure 21.5 shows the stress paths for these tests. Aslong as p > 0.5pmt the qf versus pf points fall along the

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Ch. 21 Drained Shear Strength 307

o Undisturbed soil

o Remolded soil

+ Activity > 0.75x Activity < 0.75

50 60 80 1006 8 10 15 20 30 40Plasticity index (%)

Fig. 21.4 Relationship between sin 4> and plasticity index for normally consolidated soils (From Kenney, 1959).

sarheline obtained from tests on normally consolidatedspecimens. For lower pf the points fall above the relationfrom normally consolidated tests. Thus preconsolidationaffects the effective stress-strength relation and tends tomake the sample stronger at a given pf. This preconsoli-dation effect is difficult to see when results are plotted tothe scale of Fig. 21.5a, hence the portion of this plot nearthe origin is magnified in the lower portion of the figure.

Example 21.1A specimen of Weald clay is consolidated to 100 lb/in.2, and

is then failed by decreasing 63 while al is held constant. Findqf,Pf, and wf.

Solution. On part (c) of Fig. 21.1 draw the effective stress

Qf>Pf

Fig. E21.1

path for this loading until it intersects the qf-pf relation (seeFig. E21.1). The point of intersection gives qf and pf. Thengo to Fig. 21.3, enter with either qf orpf, and read wf.

Answers. qf = 27 lb/in.2, pf = 73 lb/in.2, w, = 19.2%.Note that vr increased slightly during shear. A

80

60

,

40

20

0i

20

10

1a

il CO nso

for

normc idation

/

//

/ / /r ?

20 40 60 80

_

( i + 03)P = --

(a)

100 120 140 160

(lb/in.2)

a-

V /I J!-

vs.

/ normal co

from

isolidation

10

p = .

20

( i + 3)2

(b)

30 40

(lb/in.2)

Fig. 21.5 Results of CD tests on overconsolidated Wealdclay. pm = 120 lb/in.2

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308 PART IV SOIL WITH WATER-NO FLOW OR STEADY FLOW

20I

vs. p,

for

120 lb/in.*m

CO

10Straight line

fit

ta

0To 10 20 30 40 50

(lb/in.2)f

Fig. 21.6 Straight line fit to curved Mohr envelope for overconsolidated Wealdclay.

a = 1 lb/in. sin cfe = tan b20.3 c = a/cos

The effect of preconsolidation is seen in all of theserelations. Thus the wf versus qf and wf versus pf relationsare different for overconsolidated specimens than for thenormally consolidated specimens: for a given pf thewater content is smaller (corresponding to the increasedstrength) in the overconsolidated specimen. However,the relations for the failure condition are affected muchless by preconsolidation than is the m;0 versus Pq relation.Thus shearing action tends to destroy the effects ofpreconsolidation but does not quite succeed.

Figure 21.7 is especially useful for illustrating thechanges in water content during shear. Some of thespecimens increased in watei content during shear,whereas others decreased in water content. For example,the specimen that was rebounded to 15 lb/in.2 increasedin water content from 20.5 to 21.3%; i.e., water wassucked into the specimen during the shear part of thetest. On the other hand, the specimen that was reboundedto 60 lb/in.2 decreased in water content during shearfrom 18.9 to 18.3 %. A sample with an overconsolidationratio of about 4 would not have any net volume changeduring the shear process.

The q/pj-Wj RelationshipWe have now seen that a soil does not possess a truly

unique relationship among shear strength, effectivestress at failure

,

and water content at failure-this

relationship is affected by the degree of preconsolidation.1This will be the first of several instances in which we

shall see that the qt - pf - wf relation is only approxi-mately unique. Actually, the concept of such a relationis qualitative rather than quantitative. It reminds us thatthe three qualities are closely interrelated and helps usto envision how changes in either pf or \v} will affectstrength and each other.

Having determined this relationship for a givenpreconsolidation, we can use it to find the strength forany type of loading. Examples corresponding to thosefor normally consolidated soil will be found among theproblems.

Comparison of Clay and SandThe results presented here again emphasize the

similarities in the strength behavior of sand and clay.For both types of soils

, the shear strength is stronglydependent on the effective stress existing at the time offailure. Furthermore

, in both types of soil, the shearstrength is also dependent on the initial void ratio beforeshear, i.e., the degree to which the soil has been "pre-densified" by some past action. In a general way, densesands correspond