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MESSAGE

Geotechnical investigation is the most important component of Geotechnical Engineering practice and case studies of failures of structures on account of poor soil investigations are numerous. I am happy to note that Bhubaneswar chapter of IGS, which is one of the youngest chapters of IGS has been doing a great service to the cause of Geotechnical Engineering and the two day national event in this area is timely.

I hope that many organizations involved in soil testing will participate and it is desirable that the chapter prepares a directory of Geotechnical services ln the state of Orissa. It is expected that regular short term courses by the chapter as well as Institutes such as IIT bhubaneswar and other Institutes are conducted regularly.

I wish the two day National workshop on‘Geotechnical Investigation & Soil testing’ during 27.1.2018 & 28.1.2018 a great success.

Prof. G.L. Sibakumar Babu

President Indian Geotechnical Society

MESSAGE

I am happy to learn that Indian Geotechnical Society and School of Civil Engineering, Kalinga Institute of Industrial Technology, Deemed to be University are jointly organizing a Two days National Workshop on “Geotechnical Investigation & Soil Testing” on 27th and 28th January, 2018. The main purpose of this workshop is to refresh and enhance the knowledge base of practicing Civil Engineers on various soil testing methods and geotechnical investigation practices which are essential for planning and construction of any civil engineering structure.

The workshop will be graced by the eminent Professors and Practitioners which will provide a great opportunity for the participants to interact with the leading geotechnical experts of our country.

I wish the Workshop a great success.

Registrar

Kalinga Institute of Industrial Technology (Deemed to be University)

MESSAGE I, on behalf of School of Civil Engineering, feel highly honored to host Two Days National Workshop on "Geotechnical Investigation and Soil Testing" in association with Indian Geotechnical Society in Kalinga Institute of Industrial Technology (KIIT), Deemed to be University on 27th and 28th January 2018. The workshop will focus on Geotechnical engineering issues, remedial measures and hands on training using State of the Art laboratory facilities of School of Civil Engineering. This workshop will be largely attended by practicing engineers and consultants, engineers from Government and private organizations, academicians and students. I strongly belief that the workshop will greatly influence and benefit all its participants. I hope the opportunity to interact with the leading Geotechnical Experts in the country will be the highlight of the event.

I express my sincere gratitude to Dr. Achyuta Samanta, Honorable Founder (KIIT & KISS) and respected management authorities who are always the source of inspiration to carry out such programme at our University premises.

I extend a warm welcome to all the participants and wish the Workshop a grand success.

Dr. Benu Gopal Mohapatra

Professor & Dean School of Civil Engineering email: dean.civil@kiit.ac.in

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CONTENTS

Topic Author Page

1. IMPORTANCE OF SOIL INVESTIGATIONS IN CIVIL ENGINEERING CONSTRUCTION AND EARTHQUAKE EFFECTS

Dr. H N Ramesh 1

2. CASE STUDIES ON GEOTECHNICAL FAILURES DUE TO LACK OF PROPER SITE INVESTIGATION

Dr. C.N.V. Satyanarayana Reddy

13

3. MECHANICAL COMPACTION - A SIMPLE GROUND IMPROVEMENT TECHNIQUE: A CASE STUDY

J.C. Gogoi, Dr. A.A. Laskar

22

4. INFLUENCE OF CHEMICAL CONSTITUENTS OF LEACHATES ON THE SWELLING BEHAVIOUR OF COMPACTED BENTONITES

Jagori Dutta 30

5. SUITABILITY OF SLOPE STABILIZATION TECHNIQUES : AN ASSESSMENT BASED ON CASE STUDIES

Oindrila Das, Ansel Jose, Satya Ranjan Samal

36

6. EFFECT OF CONFINING PRESSURE AND VOID RATIOS ON SHEAR WAVE VELOCITY FOR COHESIONLESS SOIL

Dr. Rana Chattaraj, Mr. Prabhakar Kumar, Ms. Neha Nasreen

42

7. STRENGTH PARAMETER UNDER PLANE STRAIN CONDITION

Dr. Satyajeet Nanda, Dr. Benu Gopal Mahapatra Mrs. Bandita Paikaray

47

8. EMPERICAL STUDY OF AXIALY LOADED ROCK SOCKETED PILE IN BHUBANESWAR REGION

Hemanta K. Dash, Aradhana Mishra, Sitaram Satapathy

54

9. IMPORTANCE OF NABL ACCREDITATION AND PROCEDURE

Venugopal C., Er. Laxmi Kanta Tripathy

57

10. CODES AND STANDARDS RELATED TO DEEP FOUNDATIONS

62

11. APPLICATION FORM FOR FELLOWSHIP / MEMBERSHIP

66

1

IMPORTANCE OF SOIL INVESTIGATIONS IN CIVIL ENGINEERING CONSTRUCTION

AND EARTHQUAKE EFFECTS

Dr. H N Ramesh INTRODUCTION Soil as a three phase system is a well known fact. The three phase system figure is an amazing figure from which 15 definitions and 19 inter relationships can be derived. However, This geometrical figure is an assumed figure, However, the figure is practically valid which is Amazing. Howvever, in Geotechnical Engineering following points are very popular namely:

• Soil is a complex material • Soil is a three phase system .

Important types of Soils As civil engineer one should view from the construction point of view generally following types of soils are considered based on mineralogical contents namely

Montmorillonitc soils – Expansive Kaolinitic soils – Non- Expansive

Strength, Compressibility and Permeability of Soils are the three engineering properties of soil.

Behaviour of Soils : The behaviour of soils for construction purposes based on various points as follows :

• During Earth quake • Under Industrial buildings • Under wheel loads • In Case of Residential buildings or lightly loaded structures • Settlement of soils

The engineering properties of soils are having significant role in the design of foundations:

1. Compressibility 2. Strength of soils 3. Permeability of soils

2

Types of Foundations Based on the Index properties, Engineering properties and based on the load transferred from the super structure to the foundation the foudations are Generally classified as :

1. Shallow foundation 2. Deep foundation

Economy and Safety of any civil Engineering structure depends on : 1. Water table effects 2. Land filling 3. Roots of Trees – Natural Stability 4. Stability of Slope

Soil below ground level is unpredictable. It may be having silt, clay, sand, gravel etc or various combination of the above, Rock or water table may be encountered. Exploration of the soil below Ground level and analysis of soil for the suitability for the construction purpose has significant role from the economy and safety point of view.

Hence to know the Index and Engineering Properties of soil, types of soil etc., soil Investigation is necessary.

Definition-Site Investigation The process of determining the layers of natural soil deposits that will underlie a proposed structure and their physical properties is generally referred to as site investigation.

The purpose of a soil investigation program 1. Selection of the type and the depth of foundation suitable for a given structure. 2. Evaluation of the load-bearing capacity of the foundation.

3

3. Estimation of the probable settlement of a structure. 4. Determination of potential foundation problems (for example, expansive soil,

collapsible soil, sanitary landfill, and so on). 5. Establishment of ground water table. 6. Prediction of lateral earth pressure for structures like retaining walls, sheet pile

bulkheads, and braced cuts

Methods of Soil Investigation The soil investigation is necessary to analyse the soil behaviour for the construction of any civil engineering structures. However, each method will be selected based on the the type of structre, loads , type of soil etc. as follows:

1. Soil boring, Standard Penetration test. 2. Static cone penetration 3. Dynamic Cone penetration test 4. Plate load test, Field vane shear test From the field investigation following factors will be known:

1. Soil strata below nthe ground level at various depths like silt,clay, rock etc 2. Depth of water table,Rock strata 3. Undisturbed and disturbed samples can be collected at various depths below the GL 4. Strength of field soil at various depths can be determined using SPT and at Shllow

depth by plate load tests. 5. The samples collected by field investigation can be tranpored to Laboratory for tests

Laboratory tests The samples collected from the field investigations will be analysed to determine the Index and Engineering properties of the soil. The undisturbed samples will be used to determine the field water content, density, compressibility, strength and permeability. Disturbed samples collected will be used to determine the Index properties of the soil. The important Geotechnical properties to be determined in the laboratory are listed below:

1. Liquid limit, Plastic Limit, Shrinkage Limit. 2. Grain size distribution 3. Strength test 4. Permeability 5. Compaction 6. Consolidation

4

The contents of Soil Investigation Report Based on the detailed field and Laboratory investigations, Soil Investigation report will be prepared. Following salient points are to be considered

1. Site location Plan 2. Type of Structure coming up 3. Nature of soil 4. Bearing capacity of soil, settlements 5. Recommendation of type of foundation 6. Ground improvement techniques required if any

Safety In Construction Safety is a mandatory thing to the construction industry, as the human being is directly or indirectly involved in the construction activity, their life is more important compared with the luxury of the construction according to labor law of Indian Construction.

Quality In Consturction Quality in construction is grouped in to two stages 1. Quality before starting the construction. 2. Quality during and after the construction

Conclusion 1. Soil investigation is necessary for all types of structures whether it is residential

or otherwise. 2. Approval should be sought from the concerned authorities before the

construction of civil engineeringconstruction. 3. Geotechnical consultant and forensic Geotechnical engineer need to be consulted

before the construction of any structure. 4. Chemical tests and its analysis, opinions of soil chemist is necessary in any of the

structures where specialized techniques are adopted.

EARTHQUAKE EFFECTS Earthquake is a natural catastrophe which can not be predicted. However, by using the codal provisions and proper soil investigation and engineered design of foundation reduces the damge which will be caused by earthquake. In 2001 a major earth quake occurred in India, All most all damages are due to classical geotechnical failures and mainly due to Liquefaction. The case study is an eye opener for civil engineers to understand the importance of Geotechnical Investigation for civil engineering constructions.

5

Recent major Earth quakes in India 1. Sumatara Earthquake (2004) 2. Kashmir Earthquake (2005) 3. Sikkim Earthquake (2011) 4. Pakistan Earthquake (2013)

Global loss due to natural disaster

6

EARTHQUAKE Vibrations of the earth surface caused by waves originating from a source of disturbance in the earth mass Earthquake may be caused by volcanic eruption or by strain building process inside the earth mass.

Focus and Epicenter of Earthquake

Compared to most disciplines of civil engineering, Geotechnical earthquake engineering is quite young. While the damaging effects of earthquakes have been known for centuries, the strong contribution of soils to the magnitude and pattern of earthquake damage was not widely appreciated until relatively recently. Following damaging earthquakes in 1964 in Nigata, Japan and Alaska, and spurred by the growth of the nuclear power industry in the 1960s and 1970s, the field of geotechnical earthquake engineering has grown rapidly. Although much remains to be learned, the field has matured to the point where generally accepted theories and analytical procedures now exist for many important problems.

Earthquake engineering deals with the effects of earthquakes on people and their environment and with methods of reducing those effects. It is a very young discipline, many of its most important developments having occurred in the past 30 to 40 years. Earthquake engineering is a very broad field, drawing on aspects of geology, seismology, Geotechnical engineering, structural engineering, risk analysis and other technical fields. Its practice also requires consideration social, economic, and political factors. Most earthquake engineers have entered the field from structural engineering or Geotechnical engineering backgrounds, a fact that is reflected in the practice of earthquake engineering.

The study of earthquakes dates back many centuries. Written records of earthquakes in China date as far back as 3000 years. Japanese records and records from the eastern Mediterranean region go back nearly 1600 years. In the United States the historical record of earthquakes is much shorter, about 350 years. On the seismically active west coast of the United States. Earthquake records go back only about 200 years.

7

Compared with the millions of years over which earthquakes have been occurring, humankind’s experience with earthquakes is very brief.

Today, hundreds on millions of people throughout the world live with a significant risk to their lives and property from earthquakes. Billions of dollars of public infrastructure are continuously at risk of earthquake damage. The health of many local, regional and even national economies are also at risk from earthquakes. These risks are not unique to the United states. Japan, or any other country. Earthquakes are a global phenomenon and a global problem

Earthquakes have occurred for millions of years and will continue in the future as they have in the past. Some will occur in remote, undeveloped areas where damage will be negligible. Other will occur near densely populated urban areas and subject their inhabitants and the infrastructure they depend on the strong shaking. It is impossible to prevent earthquake from occurring, but it is possible to mitigate the effects of strong earthquake shaking: to reduce loss of life, injures, and damage.

Foundations • Majority of buildings in Kutch region were founded on shallow foundation. • Top 300 mm to 600 mm is made of Black Cotton Soil, below which is murram or

disintegrated rock. • Better quality construction comprises of coursed stone masonry for a depth of at

least 1 m with reinforced plinth beam • Poor quality construction puts random rubble masonry made of trap stone in top

soil only. • RCC structures are also founded on shallow isolated footings

Collapsed Buildings in Rapar City

1. Close to the epicenter 2. Mostly, buildings are of random stone masonry with poor or no mortar joints.

Heavy tiled roof 3. Most of the buildings are built without engineered design

8

4. Most Foundations consist of shallow wall footings 5. Mere earthquake force was sufficient to uproot the buildings

Why buildings failed? 1. Bad construction quality? 2. Not following building bylaws of city properly? 3. Improper design? 4. Not considering earthquake resistant design methods(Zone 3)? 5. Stress on economy (money & area)? 6. Poor material standards? 7. Liquefaction of ground?

SUMMARY • In soft alluvium, amplification of ground motion as high as 19.46 was estimated. • Estimated natural frequency of ground, the natural frequency of multi storeyed

structures and the predominant frequency of earthquake matched suggesting near resonance conditions.

• Site amplification and near resonance conditions were the two major causes for failure.

• More rational codal provisions are necessary.

Highways and Railways • Roads are classified as National Highways, State Highways, Major District

Roads, Other District Roads and Village Roads. • Over 5400 km of roads. Roads in Kutch district are damaged most. • Over 900 km length affected. Cost of repair is over 100 million USD • Damages include

– Longitudinal Crack along central carriage way & shoulders of embankments – Transverse cracks b/w bridge spans & roadway – Settlement/Uplift @ some location – Slope failure – Drainage structures across the road – concrete or masonry box culverts and pipes failed.

• Some Railway bridges were damaged • Bridge approaches settled • Lateral spreading caused track misallignment • Railway line near Navalakhi port experienced settlement of order of 3 to 4 m in

addition to lateral spreading • No damage to runway in Bhuj airport

9

Dams and Embankments

• Most classical geotechnical damage of Bhuj earthquake is the failure of almost all dams in Kutch and its neighbourhood.

• 20 Medium & 165 minor dams in Kutch (21 suffered severe damage, rest 164 suffered moderate to major damages)

Port Facilities

Navalakhi Port – Ground Cracks

Railway track near Navalakhi Port

Highways and Railways….

10

Summary

• Both Kandla port and Navalakhi port suffered damages • Damage in Kandla port included failure of pile foundation supporting berths 1 to

5, tilting of Port and Customs office • Damage in Navalakhi port consisted of lateral movement of walls, subsidence,

ground movement, ground break, etc.

2

1

8 m

Navalakhi Port

Lateral Spread > 10m

Subsidence > 3 m

Tilt & Settlement

Pitching disturbed

Arabian

Sea

Quay

Wall

Arabian Sea

Sheet Pile Wall

moved

Failed Bulkhead

(Wharf)

Navalakhi Port

Ground Movement

Lateral Spread > 10 m

Slope Failure

Buildings failed

Railway track

dislocated

Large subsidence >3m

11

Codal provisions in India • Codal provisions in earthquake engineering in India exist since 1950s • Only 3 or 4 modifications incorporated • Codes in masonry and structural engineering fairly sufficient • Codes on embankments, retaining walls and bridges are being improved • Design recommendation is based on strength criteria, not on permissible

deformation criteria

Local Report • Large column of water upto 6 m for over 30 minutes beyond earthquake as hot

as tea was reported. • Normally saline ground water turned sweet. Same report was made about 1819

earthquake also. • Chang dam was blasted and breaching was sudden.

Concluding Remarks • Large lengthy faults were not visible at the ground surface • Geotechnical failures did not deviate from the routine types. • Liquefaction was reported mainly by formation of craters, sand blows and lateral

spreading. • Flow failures and landslides were minimal. • Most dramatic geotechnical failure was the damage to nearly all earthen dams in

the region. Subsoil failure was the major cause. • All dams were almost empty following two seasons of draught. Otherwise

damage would have been more severe. • Surajbari highway bridge was the only major bridge to suffer. A few old and

small bridges failed. • Most buildings in Kutch were built on shallow foundations. Buildings along

with foundations were up rooted by the mere earthquake force because of closeness to epicenter.

• Failure of a few buildings in Ahmedabad is a classical geotechnical problem of amplification of ground motion and near resonance condition, perhaps in addition to other causes.

• Post earthquake data collection, particularly related soil behaviour and geotechnical facilities needs improvement.

12

• IS Codes are still being improved. While design of embankments and dams is still based on strength, present day understanding of design from the consideration of allowable settlement needs to be practiced.

• Microzonation of important cities, particularly in seismically active zones is useful.

• In general, Bhuj earthquake was an eye opener to bureaucrats, politicians, general public, technologists and researchers in India

• Engineered structures faired better suggesting the importance of adopting strict seismic design in the areas prone to earthquake.

Aspects of Earthquake geotechnical Engineering • Determination of dynamic loading from anticipated earthquake • Field and laboratory tests to determine soil modulus and damping characteristics

of soil • Estimation of Liquefaction Resistance and identifying liquefaction prone areas • Estimation of settlement of soil (dry/partially/fully saturated) and structures • Analysis of Bearing Capacity, Slope stability and Retaining Walls for

anticipated earthquake • Other earthquake effects such as surface rupture of roads, pavements and

embankments, pipe line design etc. • Site improvement methods and foundation alternatives • Building Code Provisions pertaining to EGE specifications of the region

13

CASE STUDIES ON

GEOTECHNICAL FAILURES DUE TO LACK OF PROPER SITE INVESTIGATION

Dr. C.N.V. Satyanarayana Reddy1

INTRODUCTION

Though soil investigation importance is realized over the past two to three decades, still the way in which it is being done is resulting in failure of structures. Soil investigations are essential in design of foundations of structures and in selection of soils from borrow areas with intended use as constructional materials. Soil Investigation programme will be complete and useful only if it is done addressing the issues of purpose, extent and depth of Investigation. Expertise of geotechnical expert and skill of persons involved in testing play a key role in obtaining reliable geotechnical investigation report. In case of pile foundations, it is essential to verify the estimated pile load capacities from load tests.

The paper presents two case studies to illustrate the importance of proper site investigation to avoid failures in the structure being built and also in existing neighboring structures. The first case study deals with affected railway track due to ground heaving adjacent to dumping yard in Visakhapatnam port area while the other case study deals with excessive settlement of a pile group supporting one of the trestles of conveyor belt. The case studies emphasis

CASE STUDY 1 : RAILWAY TRACK AFFECTED BY GROUND HEAVING

Major Ground heaving problem occurred at a stacking yard in Visakhapatnam Port area in February 2008 affecting railway line (being used by Port) that was passing by the side of the yard. The failure created tension in the public as they suspected it due to an Earthquake or as warning of a probable severe earthquake. News papers published the event as a Land Tsunami. Based on the direction of district administration, the Geotechnical Engineering Division of Andhra University investigated in to the failure and clarified that it is a common phenomenon that occurs in clays of soft to medium consistency when stress due to loading exceeds net ultimate bearing capacity of supporting soil. The details of the case study are presented below. 1 Professor, Geotechnical Engineering Division, Department of Civil Engineering, College of Engineering, Andhra University, Visakhapatnam – 530003, Andhra Pradesh E mail: cnvsreddy@gmail.com

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14

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15

Manganese Ore The stacked manganese ore in port area at stack yard has been tested in the laboratory and its properties are presented in Table 2.

Table 1. Engineering Properties of Marine Clay S. No. Engineering Property Value

1 Specific Gravity 2.48 2 Grain Size Distribution

a) Gravel (%) b) Sand (%) c) Fines (%)

0 12 88

3 Plasticity Characteristics a) Liquid Limit (%) b) Plastic Limit (%)

65 32

4 IS Classification CH 5 Free Swell Index 77 6 In-Situ Density (g/cc) 1.75 7 Natural Water Content (%) 56.8 8 Undrained Cohesion (t/m2) 1.8 9 Net Ultimate bearing capacity of soil (t/m2) 6.2

Table 2. Properties of Manganese Ore S. No Engineering Property Value

1. Specific Gravity 4.47 2. Grain Size Analysis

a) Gravel (%) b) Sand (%) c) Fines (%)

26 64 10

3. Dry Density (g/cc) a) Loose State b) Dense State

2.58 3.20

Taking unit weight of manganese ore as 2.58 t/m3 and considering the stack load as rigid, the average vertical stress induced at marine clay strata surface at a depth of 1.5m below ground level has been worked out as 12.9 t/m2 based on 2:1 load dispersion. The induced stress is much higher than net ultimate bearing capacity and hence, local shear failure occured involving plastic flow of soft marine clay in large volume. This has resulted in heaving of adjacent ground and there by the existing railway track got lifted up and distorted. Excessive ground heaving also led to splitting of ground surface as soil from loaded area tried to escape from ground surface.

16

Subsequent soil investigations revealed presence of marine clay with N values of 2 to 4 upto 18m depth, underlain by soft disintegrated rock. Based on Prandtl’s theory (1920), the heaving occurs to a distance of about 0.707B from boundaries of loaded footing on either side, where “B’ is width of footing as shear zones extend laterally to that distance. As base width of stacking is 32 m, the heaving should theoretically extend to a distance of 22.6m whereas the extent of heaving on ground surface at study area was observed to be upto lateral distance of 15-20m. The shear movements in foundation soil extend to 0.5B below loaded area, i.e., 16m. As the area is comprised of marine clay up to 18m, the loading from manganese ore stacking has resulted in large movement of soil vertically and laterally, resulting in upheaval and splitting of ground adjacent to loaded area. The maximum upheaval is noticed at about 12 -16m distance from the area of stacking. As the railway track is located at a distance of 12m from stacking embankment, it got pushed up due to underground plastic flow of soil and experienced severe damage.

Conclusion The ground heave and damage to railway track have occurred due to lack of proper soil investigation. Proper soil investigation would have enabled either restricting the stacking height of the material or planning ground improvement programme for strengthening of the soft clay.

CASE STUDY 2 - SETTLEMENT FAILURE OF PILE GROUP One of the Material Handling Firm at Visakhapatnam Port has planned for new storage facility by laying new conveyor belt. To support the conveyor belt, Two RCC Columns supported on a group of four piles have been erected and horizontal frame work for supporting conveyor belt was laid. Four Precast concrete piles of 450mm diameter and 15.5m length were used based on soil investigation report which they got from a private agency. It is considered from the soil investigation report that rock is present at 15.5m depth and piles are terminated. While executing the work, the columns supporting the frame work of conveyor belt settled by about 100mm due to self weight as shown in Fig. 1, even without placing conveyor belt. The settlement resulted from piles supporting the columns.

The Material Handling Firm approached Andhra University to investigate in to the problem of settlement of piles and to suggest remedial measures. Based on site visit, it is observed that the area is dumped by coke up to 4.0m high over a large extent. It is informed that the problem of pile settlement started after dumping of material in that area. To investigate into the failure, fresh soil investigation was done.

17

Details of Investigation The site investigation was done with borehole of 150mm diameter and 26.5m length at the affected area. The bore log details and engineering properties of subsoil strata are presented in Table 3 and Table 4.

Table 3. Bore Log of Affected Pile Group Location Dates of Exploration: 22-07-2012 to 24-07-2012

Location : Visakhapatnam Port Depth of GWT: 2.3m

Bore Hole No : BH-1 Depth of Bore Hole : 26.5m Type of Boring : Rotary Diameter of Boring : 150mm

Depth (m) Description Type Depth

(m)

Blow count for split spoon sampler penetration through N CR

(%) RQD(%) 0-15cm 15-30cm 30-45cm

0.0-0.5 Filled up soil (Clayey sand with gravel)

DS 0.0 0.5

0.5 -10.0

soft marine clay

DS 0.5 3.0 SPT 3.0 3.45 01 01 02 03 DS 3.5 7.5 SPT 7.5 7.95 02 02 02 04 DS 8.0 10.0

10.0-13.5

Clayey gravel

SPT 10.5 10.95 09 14 23 37 DS 11.0 13.0 SPT 13.0 13.40 30 42 Refusal >100

13.5-19.0

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DS 13.5 14.5 SPT 14.5 14.54 Refusal - - >100 SPT 16.0 16.45 24 30 35 65 SPT 17.5 17.58 Refusal - - >100 DS 17.58 19.0

19.0-22.0

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Rock

SPT 19.0 19.01 Refusal - - >100

RCS 19.0 22.0 26 0

22.0-22.5

Soft Disintegrated Rock (SDR)

DS 22.0 22.5

22.5-26.5

Fractured Hard Rock

RCS 22.5 24.5 26 10 RCS 24.5 26.5 33 14

Bore termination Depth:26.5

Notations: DS : Disturbed Sample SPT : Standard Penetration Test RCS : Rock Core Sample N : Standard Penetration Resistance CR : Core Recovery RQD : Rock Quality Designation

18

The sub soil was found to be consisted of filled up soil of 0.5m thickness, soft marine clay of 9.5m thickness, clayey gravel of 3.5m thickness, SDR of 5.5m thickness, Highly fractured rock, SDR followed by Fractured hard rock. The depth where the pile was terminated (i.e. 15.5m) consisted of soft disintegrated rock (SDR) of varying stiffness. The ground water table was present at a depth of 2.3m below ground surface at borehole location.

Table 4.Engineering Properties of Soil at Affected Area

Depth (m) Subsoil strata ISC N

Particle size Analysis

Plasticity characteristics ρ

(g/cc)NMC (%)

FSI (%)

C (t/m2)

Ø (Deg)G

(%)S

(%)F

(%)LL (%)

PL (%)

PI (%)

0.0-0.5 Filled up soil (clayey sand with gravel)

SC -- 20 38 42 34 21 13 - - - - -

0.5-10.0 soft marine clay

0.5-3.0 -do- CH -- 00 32 68 60 28 32 - - - - - 3.0-5.0 -do- CH 03 01 30 69 63 32 31 2.02 65.5 60 1.6 12 5.0-7.5 -do- CH --- 00 25 75 58 28 30 - - - - -

7.5-10.0 -do- CH 04 00 22 78 60 28 32 2.01 61.6 50 1.8 10

10.0-13.5 Brown clayey gravel

10.5-10.95 -do- GC 37 43 32 25 43 24 19 2.28 25.8 35 1.2 29

13.0-13.40 -do- GC >100 36 40 24 42 24 18 2.27 24.6 - - -

13.5-19.0 SDR 13.5-14.5 -do- - -- 29 48 23 32 20 12 2.22 20.7 - - -

14.5-14.54 -do- - >100 Sample inadequate for analysis

16.0-16.45 -do- - 65 20 52 28 31 19 12 2.20 15.81 10 - -

17.5-17.58 -do- - >100 No sample recovered

17.58-19.0 -do- - -- 25 55 20 28 19 09 - - - - -

19.0-22.0 Highly

Fractured Rock

19.0-19.01 -do- --- >100 No sample recovered

19.0-22.0 -do- --- --- Rock core sample recovered with CR 26.66% and RQD NIL 22.0-22.5 SDR - ---

22.0-22.5 -do- - --- 59 13 28 30 19 11 - 7.54 - - -

22.5-26.5 Fractured hard rock

22.5-24.5 -do- --- Rock core sample recovered with CR 24.4% and RQD 10.5% (compressive strength of rock = 3210t/m2)

24.5-26.5 -do- --- Rock core sample recovered with CR 32.8% and RQD 14.6% (compressive strength of rock =3450t/m2)

19

Notations: G : Gravel ISC : Indian Standard Soil Classification Symbol S : Sand ρ : In-situ Density F : Fines C : Cohesion LL : Liquid Limit Φ : Angle of Internal Friction PL : Plastic Limit PI : Plasticity Index NMC :Natural Moisture Content FSI : Differential Free Swell

Based on the bore log information, the allowable load capacity of pile is determined as per IRC 78-2014 and IS 2911 part 1 (section 2) to arrive at the cause for the settlement of piles of conveyor belt supporting system.

Piles terminated in SDR

The ultimate end bearing resistance = Cub × Nc where, Cub = shear strength of soft disintegrated rock Nc = 9.0

Allowable end bearing resistance is taken as ultimate bearing capacity divided by a factor of safety of 3.0

Ultimate skin friction resistance (side socket shear) in MPa = 0.225√qc

Allowable skin friction resistance is taken as1/6th of ultimate skin friction resistance.

Piles terminated in Fractured Hard Rock:

The ultimate end bearing resistance = qckspdf

where, qc = uniaxial compressive strength of rock df = depth factor = 1+0.4 ls/D = 1.4(for a socket length of D) …. limited to 1.2; ksp= Factor accounting for discontinuities in rock {taken as 0.3 as (CR+RQD)/2 is about 30%

Allowable end bearing resistance, qs is determined as ultimate bearing capacity divided by a factor of safety of 3.0.

Ultimate skin friction resistance (side socket shear) in MPa = 0.225√qc

Allowable skin friction resistance, qs is taken as1/6th of ultimate skin friction resistance.

By limiting the corrected “N” value to 60 in SDR strata, the allowable end bearing and side socket shear resistances are determined as 126t/m2 and 3.35t/m2 respectively. The ultimate skin friction resistance (kN/m2) in clayey gravel is taken as “2N”. In marine clay, the ultimate skin friction resistance is determined as 0.9 times undrained

20

cohesion of soil. Undrained cohesion (kg/cm2) of clay is taken as 1/16th of ‘N’ value. The ultimate skin friction resistance in clayey gravel and soft marine sandy clay has been taken as 7.4t/m2 and -1.7t/m2 respectively. A factor of safety of 2.5 has been used to arrive at allowable skin friction resistance in clayey gravel. The negative skin friction effect of marine clay has been considered in evaluation of load capacity of pile. The load capacity of pile is estimated considering provision for permanent liner in top 1.5m depth.

Load capacity of pile is estimated using static formula as

Allowable Load capacity of pile, Qa = Qs + Qeb = ∑ fs. As + Ap.feb

where, Qs is ultimate capacity in skin friction and Qeb is ultimate capacity in end bearing; As is surface area of pile in component layer; fs is average skin friction resistance in component layer ; feb is end bearing resistance, Ap is cross sectional area of pile at tip.

The pile load capacity of 450mm diameter pile of 15.5m length has been assessed based on soil investigation data as detailed below.

Table 5. Details of load capacity estimation of affected pile

S.No. Soil layer Thickness (m) Allowable load

capacity in skin friction (t)

Allowable load capacity in end

bearing (t)

1 Soft marine clay 9.0 -21.6 ---

2 Clayey gravel 3.5 14.7 ---

3 SDR 2.0 9.5 20.0

Based on the details presented in Table 5, the allowable load capacity after accounting for downward drag from soft marine clay was determined as 22.6t and was found to be much less than design load capacity of pile reported to b e 65t). Since dumping of material was done at the area, the downward drag force developed on the pile surface and resulted in reduced allowable load capacity and led to settlement of piles.

In view of the prevailing situation, it was advised to install new piles to support the supporting columns of conveyor belt by terminating the piles in fractured hard rock available at 22.5m by maintaining a minimum socketing length of ‘d’, where ‘d’ is the diameter of pile. The allowable vertical Compression load Capacity of 600mm dia. and 23.5 m length pile at location represented by Bore Hole is estimated as 103t. Side socket shear resistance is considered over a length of 6d from pile termination level in evaluation of load capacity of piles.

21

Conclusion: Proper soil investigation would have avoided the settlement failure of pile group. Also, if any pile is laid at a location which is not thoroughly investigated, routine pile load tests shall be conducted to check the adequacy of proposed pile to support the intended design load.

REFERENCES

1. Bowles, J.E. (1997). Foundation Analysis & Design, 5th Edition, Mc Graw Hill Publishing Co., New York.

2. Hausmann, M. R. (1990). Engineering Principles of Ground Modification, Mc Graw Hill Publishing Co., NewYork.

3. IS: 6403-1981, Code of Practice for Determination of Bearing Capacity of Shallow Foundations, BIS.

4. IRC 78-2014: Standard specification and code of practice for road bridges Section-VII (foundations and substructure).

5. IS 2911 part 1 (section 2) -2010: Code of practice for design and Construction of Pile Foundations, Concrete Piles- Bored cast in-situ piles, Bureau of Indian Standards, New Delhi.

22

MECHANICAL COMPACTION - A SIMPLE GROUND IMPROVEMENT TECHNIQUE: A CASE STUDY

(ISSN 2278 –0 5450) (EISSN 2278-05450)

J.C. Gogoi1, Dr. A.A. Laskar2

ABSTRACT: A mega water (208 MLD) supply project was undertaken by Guwahati Metropolitan Development Authority (GMDA) funded by Japan International Cooperation Agency (JICA) for central Guwahati region. In the detailed project report (DPR), provision of pile foundation with raft was accommodated for construction of clear water reservoir (CWR) at Kharghuli Hills at the bank of river Brahmaputra in Guwahati city. The foundation soil has been considered unsuitable for shallow open foundation due to inadequate soil pressures. The safe soil pressure at the proposed foundation depth was worked out as 7085kg/m2. Therefore, initially pile foundation was proposed with a provision of 900 bored cast-in-situ RCC piles of varying dimensions with an estimated cost of about 50 million requiring more than 12 months. Due to constrain in time and cost, an alternative foundation system became essential to complete the foundation work in less time. After a detailed feasibility study of different alternatives considering safety and economy, a simple technique of ground improvement by mechanical compaction of foundation soil at in-situ condition using road roller was considered suitable. Dynamic plate load tests were conducted at different location after compaction at in-situ soil moisture and the safe soil pressure (SBC) was worked out as 15750kg/m2 which is more than double of original SBC which was considered adequate for construction of proposed foundation. However, in addition to this, two layers of well graded granular material of 250mm thick were also provided for quick dissipate of excess soil moisture during compaction. The dynamic plate test results shows very high increase of SBC up to an extent of 46130kg/m2. The employment of this simple means save lot of money and the entire foundation works had been completed within a short time of merely of three months.

Keywords : Ground Improvement, Mechanical Compaction, Compacted Granular Sub-grade

1 Foundation Research Institute, Beltola College Road, Guwahati, India, Email: jcgogoi@gmail.com 2 Foundation Research Institute, Beltola College Road, Guwahati, India, Email: laskar.alam@yahoo.co.in

23

1. INTRODUCTION The Guwahati Metropolitan Development Authority (GMDA), Guwahati of Govt of Assam, put forward a proposal to construct a mega water supply scheme with an estimated capacity of 208 MLD to meet up crisis of potable water of citizen of South-Central Guwahati. This is ever biggest water supply scheme of North East India. The project is funded by Japan International Cooperation Agency (JICA). Louis Berger Group INC, USA has been appointed as Project Monitoring Consultant (PMC). A clear water reservoir of 42.20m×46.00m (including sump well) is to be constructed and this reservoir site of the project is located at Kharghuli hill top near river Brahmaputra (N 26°11'53.97" E 91°46'6.20").

Initially, the detailed project report (DPR) was prepared considering a provision of composite foundation system of RCC raft and about 900 bored cast-in-situ RCC piles of different dimensions (length varying from 12.00m to 15.00 m below RCC raft and diameter are in the range of 500mm and 750 mm). The estimated cost of pile foundation worked out as 50 million. The pile foundation may require about 12 months time where as JICA has a strict time frame of completion of the project within 32 months. In case of pile foundation, the project cannot be completed within the stipulated time frame. Besides, the accommodation of piles at uniform (3 diameter) pile spacing is considered difficult. As such, a feasibility study of alternative foundation system was conducted keeping view of both time and cost without compromising with the safety of structure.

2. STUDY DETAILS The foundation soil predominantly consisted of low-plastic sandy clay (CL) in normally consolidated state of 4.5m below the foundation level of RL 49.75m followed by dense and compact hill sand deposit of rock weathering origin of reasonable thickness. A few dynamic plate load tests (as per ASTM E 2835–11 by interpretation of the principles of Kinetic energy) were conducted at the proposed foundation depth of RL 49.75m to worked out the net safe bearing capacity (SBC) of the foundation soil at in-situ condition. 4 (four) open trial pit were made for collection of undisturbed soil sample to determine Specific gravity (GO), in-place moisture content (WO), density (Y) and void index (eO). The average compression index (CC) was calculated using empirical relationship of void index and compression index [CC = 0.30 (eO – 0.27)] as per IS 8009 part 1. A few bulk soil samples also collected for determination of maximum dry density (MDD) at in-place moisture content of the foundation soil since the soil moisture of the foundation soil of 4.50m thick was practically difficult to bring to optimum moisture level (OMC) (Owens and Malisch, 1989). Proctor compaction

tests (Figu

bef

aft

propan anSBCcomp

Fig

were also cure 1). The

Table 1

fore compac

ter compact

The averaosed foundanticipated fo of 15000kpaction pote

ure 2: A. B. I

conducted tfollowing p

F

1: Moisture

G

ction 2.6

tion 2.6

age net saation level o

foundation skg/m2. The entialities h

A

SBC Test (In-situ Soil w

to compare parameters o

Figure 1: Pr

e Density Re

GO Y(T/M

665 1.62

665 1.72

fe soil preof RL 49.75settlement o

proctor coave noticed

Dynamic Pwas Compa

24

the OMC of in-situ co

roctor Comp

elationship

M3) WO (%)

25 21.03

20 21.03

essure (SBC5m over theof 69.62mmompaction td. The param

late Load Tacted at In-s

and In-situondition wer

paction Curv

of Virgin an

eO

0.640

0.549

C) worked e in-situ vir

m which is mtests were cmeters are d

Test) at Founsitu Field M

soil moisture worked o

ve

nd Compac

CC

0.111

0.084

out as 70gin soil (Fig

much less thconducted a

depicted in T

B

ndation DepMoisture Con

ure content out.

cted Soil

MDD (T/M3)

O

1.782 1

085kg/m2 agure 2A) aghan designeand encourTable 1.

pth) RL 49.7ntent

(WO)

OMC (%)

18.55

at the gainst ed net aging

75m

foundused Durgcondover adequfound

Figur

optimexpecconsidownmois

On the stdation soil which prod

gunoglu et.aducted (Figur

the compacuate for thdation. The

re 3: A. SBC

The in-situmum moistucted to be aidered reasongraded cruture during

Figu

trength of cwas compa

duces lateralal., 2003) anre 3A) and t

cted soil agahe proposeddetails of sa

A C Test on Co

u soil moisure content

absorb by agonable to proushed aggrecompaction

A ure 4: A. SB

B. SB

compaction acted using l soil heavend the dynathe average

ainst an anticd raft foundample calcul

ompact In-sit

sture content (OMC), ggregate’s wovided 2 laygates + 20%

n and to abso

BC Test on BC Test on

25

test resulta road rolle) in fair we

amic plate lnet safe bea

cipated foundation whiclation presen

tu Soil. B. Co

nt appears tthis nomin

water absorptyers of 250 % crusher

orb soil mois

250mm Co500mm Gra

s at in-situer (Figure 2eather (Revilload tests ataring capacitndation settlech is adequnted for read

ompaction o

to be althounal differenction capabilmm thick odust) to qusture by abo

ompact Grananular Sub-

u soil moist2B) (initiallyl et.al., 2002t random fety worked ouement of 23uate for elimdy reference

B f 250mm Gr

ugh slightlyce (21.03–1ity (2-4%) a

of granular suick dissipatout 2% (Figu

B nular Sub-g-grade

ture conteny vibratory 2; Lersow, ew locationsut as 15742k.493mm whmination of.

ranular Sub-g

y higher tha18.55 = 2.and it is thersub-grade (4te of exces

ure 3B).

grade,

nt, the roller 2001; s was kg/m2 hich is f pile

grade

an the .48%) refore 40mm s soil

26

After providing first layer of compacted granular sub-grade of 250mm thick, a few dynamic plate load tests were conducted and average SBC was worked out as 36376kg/m2 which is about 5 times higher than the SBC of in-situ virgin soil and 2.3 times higher than the compacted in-situ soil (Figure 4A). After completion of initial layer of granular sub-grade, the final layer of 250mm was provided with proper compaction using road roller. Random dynamic plate load tests were conducted and the average net SBC was worked out as 46131kg/m2 which is 6.5 times higher than the SBC of in-situ virgin soil and 2.9 times of compacted in-situ soil (Figure 4B). The details of improvement of SBC depicted in Table 2 and in Figure 5.

Table 2: Details of SBC Test Conducted During Various Conditions

Formation Net SBC (kg/m2) Settlement (mm) Average SBC (kg/m2)

In-situ soil before compaction

7370 55.70

7085 9935 41.35 6622 61.96 4379 93.52

In-situ soil after roller compaction

15341 26.81

15742 18363 22.40 13504 30.45 15759 26.10

250 mm thick compacted

granular sub-grade

41210 52.45

36376

40347 53.57 43052 50.20 25947 83.26 33799 63.94 33898 63.75

500 mm thick compacted

granular sub-grade

54397 39.74

46131

36696 58.89 34454 62.72 44717 48.33 51261 42.17 46610 46.37 54016 40.02 46894 46.09

adop8009for d(ES) foundcorrepeakcomp

was estabfor 9Summ

R

15

GrSu

The net sapting a safet9 part I, igndesign load o

of 249236dation basection was k monsoon. pacted form

Cost analyestimated

blishment co9 months abmarised cos

RCC

50

ranular ub-grade

afe bearing ty factor ofnoring depthof 15000kg62.06 kPa e (42.20mmade sinceRigidity fa

mation and s

ysis of comas Rs. 1.7

ost even anbout Rs. 0.st estimation

Figure 5:

capacity (Sf 6.0. The sh factor mo/m2 worked

and averamx46.60m)

e the foundactor of 0.8ub-grade ar

Figure 6:

mpaction of 708 millionnd a net sav01 per litren shown be

27

Improveme

SBC) workeettlement w

odification. d out as 2.84age Poisonusing equa

dation is clo80 was adopre depicted

Compacted

in-situ soiln which isving of 96.6e which is elow.

ent of SBC

ed from dynwas calculat

The anticip4mm adoptin ratio of ation as stose to the pted in settin Figure 6.

d sub-grade

l and 500ms less than6% besides an addition

namic plateted as per rupated founding average0.500 con

tated earliesub-soil watlement ana.

mm thick gran 9 month

gain of wanal windfall

e load test reule 9.2.3.2

dation settlee Elastic monsidering ser. Water ater level dalysis. Deta

anular sub-hs (time saater supply l for the pr

RL 50.75m

RL

RL

esults of IS

ement odulus quare table

during ails of

grade aving)

tariff roject.

m

28

1. Cost of additional excavation of 500 mm for granular sub: (48m×46m×0.50m = 1104m3 @ Rs.130.00/m3) = Rs.137,280.00

2. Cost of coarse aggregates with 25% voids including cost of laying at site and levelling complete for (48m×46m×0.50m×0.25 = 1320m3 @ Rs. 955.00/m3) = Rs. 1,260,600.00

3. Cost of crusher dust 20% including laying and levelling 330m3 @ Rs. 450/m3 = Rs. 148,500.00

4. Hire charge of 18 Ton road Roller including POL for 90 days @ Rs. 1800/day = Rs. 162,000.00

Total : = Rs. 1,708,380.00

Less establishment cost of 12 – 3 = 9 month @ Rs. 2,85,000.00/month = Rs. 5,265,000.00

Additional gain by tariff of water supply for 9 month @ 208,000 per month = Rs.1,872,000.00

3. CONCLUSION Low plastic sandy-silty clay (CL) is suitable for mechanical compaction as ground improvement alternative. Mechanical compaction is a very simple ground improvement technique since no hi-tech expertise and sophisticated modern machinery necessary. In developing region, this method can easily be adopted for any type of structure at a very reasonable cost and less time. Time is always important to minimise the project cost to earn more from quick productivity.

Use of compacted granular sub-grade has an important role of draining out excess soil moisture during compaction besides absorption of sufficient moisture by the aggregates up to 2-3% depending upon the water absorption capacity of aggregates. Further coarse aggregates penetrated inside the clay transforming into mud-concrete for which higher SBC often attained.

Light weight dynamic plate load test is suitable for determination of SBC on compacted soil within a very short time. The safety factor of 6 is most appropriate for estimation of SBC since the dynamic test results often reflect higher SBC than that of static load test results. The foundation settlement preferably be estimated out using rule 9.2.3.2 of IS 8009 pt I which is very close to the static load test results.

Replacement weak soil with compacted granular sub-grade is considered most best suited techno-economical proposition which is about 66% cheaper and very less time consuming. In case of highly over-saturated fat plastic clay (φ=0), the mechanical compaction may also feasible with some additional treatment to minimise the soil moisture.

29

ACKNOWLEDGEMENT We are very much grateful to The Louis Berger Group Inc, USA for giving us the opportunity for conducting the feasibility study. We are also very much thankful to Mr M. Saikia, Secretary, Foundation Research Institute for his all time assistance and active cooperation for conducting the feasibility study.

REFERENCES

[1] Revil, A., Grauls, D, and Bre´vart, O. (2002) Mechanical compaction of sand/clay mixtures, J. Geophys. Res., 107(B11), 2293, doi: 10.1029/2001JB000318.

[2] Lersow, M (2001) Deep soil compaction as a method of ground improvement and to stabilization of waste and slope with danger of liquefaction, determining modulus of deformation and shear strength parameters of loose rock. Waste Management, 21, 161-174.

[3] Durgunoglu, H.T., Varaksin, S., Briet, S. and Karadayilar, T. (2003) A case study on soil improvement with heavy dynamic compaction, Proc. XIII ECSMGE, vol. 1, 651-656.

[4] Owens F. and Malisch, W. R. (1989) Soil Compaction in residential construction, Publication# 890855, The Aberdeen Group.

30

INFLUENCE OF CHEMICAL CONSTITUENTS OF LEACHATES ON THE SWELLING BEHAVIOUR OF

COMPACTED BENTONITES Jagori Dutta1

ABSTRACT: Landfilling has been the most widely accepted method of solid waste disposal in the various countries around the world. Compacted bentonites are candidate materials for engineered barriers for waste containment facilities. The expansive behaviour of bentonites when in contact with fluids is desirable in many waste containment applications as soil barriers in landfill liner and cover systems. The swelling and the hydraulic behaviour of bentonites are the key factors in the design of landfill liners since leachate from a waste disposal site influence these soil properties. Chemicals in the landfill leachate with low dielectric constant, high electrolyte concentration, or high cation valence may cause the diffuse double layer of bentonite to shrink which in turn leads to a change in the swelling characteristics of bentonite(Olson and Mesri, 1970; Mishra et al., 2005).Hence, in order to design a clay liner it is quite essential to study the swelling behaviour of bentonite in the presence of various chemicals present in the leachate. This study focuses on the significant influence of salts and heavy metals on swelling behaviour of compacted bentonites. The work investigates the influence of inorganic salts and heavy metals on the various stages of swelling of bentonites of different quality (varying in their liquid limit, cation exchange capacity, specific surface area, exchangeable sodium percentage). From the time swelling plots, it was observed that with the increase in inorganic salt and heavy metal ion concentrations, the time taken for the primary swelling reduces. A comparison between the two bentonites showed that at the same elapsed time, the percentage of swelling was higher for the high quality bentonite (with a higher liquid limit, CEC and SSA) in comparison to the low quality bentonite in presence of the various chemicals.

Keywords: Bentonite, heavy metals, salts, swelling, leachates, landfill liner

Introduction Expansive soils have a tendency to swell when in contact with water which makes it desirable in waste containment applications. Due to its high swelling capacity and low hydraulic conductivity, bentonite is widely used as a liner material at the waste disposal site (Daniel, 1984). Bentonite is a naturally available very highly plastic swelling clay produced by deposition and alteration of volcanic ash (Mitchell and Soga, 2005). The swelling capacity of bentonite, which in turn controls its hydraulic and compressibility behaviour, depends upon various physico-chemical and mineralogical factors. Bentonite primarily consists of a mineral called montmorillonite (Mitchell and Soga, 2005) and when it interacts with water, it forms diffuse double layer resulting in swelling of the bentonite (Norrish, 1954; Norrish and Quirk, 1954; Madsen and Vonmoos, 1989). However smectite clays like bentonites may undergo large interlayer shrinkage in contact with chemicals. The chemicals present in leachates affect the pore-fluid chemistry and controls the diffuse double layer thickness of the clay particles. Thus the swelling behaviour of bentonite when exposed to chemical constituents in pore fluids needs investigation.

Dakshanamurthy (1978) noticed two stages of swelling of clays. In the first stage of hydration of dry clay particles, water is adsorbed in successive monolayers on the surface of montmorillonite clay apart which is referred to as interlayer or intercrystalline swelling. The second phase of swelling is due to

1 Assistant Professor, School of Civil Engineering, KIIT University, Bhubaneswar – 751024 E-mail : jagori.duttafce@kiit.ac.in

31

double-layer repulsion. Large volume changes accompany this stage of swelling. The swelling curve can be divided into three phases. Initial swelling is generally less than 10% of the total swelling. This is essentially due to swelling of the bentonite clay particles within the voids of the coarser non-swelling fraction. Primary swelling develops when the void can no longer accommodate further clay particle swelling. It occurs at a faster rate. After the primary swelling was completed, slow continued swelling occurs. Sivapullaiah et al. (1996) studied on the swelling behaviour of mixtures of bentonite clay and non-swelling coarser fractions of different sizes and shapes and revealed that observed swelling occurs only after the voids of the non-swelling particles are filled up with swollen clay particles. After the intervoid swelling, primary swelling takes place which follows a rectangular hyperbolic relationship with time. Time-swelling relationships showed that swelling continued to occur for a long time after primary swelling known as secondary swelling. Lu et al.(2013) studied on the swelling properties of compacted clay and observed that the swelling process of the compacted clay substantially shows a ‘S’ type curve with time. The molding water content and soil compaction degree had significant effects on swelling and shrinkage of the clay.

The present work was carried out to investigate the swelling behaviour with respect to elapsed time, of bentonites of different quality (varying in their liquid limit, cation exchange capacity, specific surface area, exchangeable sodium percentage) in the presence of inorganic salts and heavy metals present in leachates. Because bentonites with different mineralogical composition may behave differently in the presence of various contaminants, it is quite essential to compare the swelling behaviour of different quality bentonites in the presence of the contaminants. The time swelling relationship of two bentonites varying in their mineralogical composition has been studied in the presence of inorganic salts and heavy metals of various concentrations.

2. Materials and Methods

2.1 Bentonite Two bentonites of different mineralogical composition and swelling properties used for the studies were procured from Rajasthan state of India. These bentonites are named as Bentonite-A and Bentonite-B in the further discussion for brevity. The physical and chemical properties of the bentonites are listed in Table 1. The data in Table 1 shows that the Bentonite-B, which has a higher liquid limit, plasticity index, clay content, cation exchange capacity (CEC), exchangeable sodium percentage (ESP), and SSA, swells more in comparison to Bentonite-A and termed as a high quality bentonite (Mishra et al., 2011). The data from the X-ray diffraction (XRD) test showed that the Bentonite-A was consisting of 59% of mineral montmorillonite; whereas, Bentonite-B was consisting of 78%.

2.2 Various chemicals used for the study For inorganic salts, solutions of NaCl and CaCl2 were chosen for this study since the leachate of the fly ash and bottom ash, which are dumped in the landfill site, mostly contains ions of Na+ and Ca2+

(Ohtsubo et al., 2004). Solutions of 0 (i.e. De-ionized (DI) water), 0.01, 0.1 and 1N concentrations were prepared by dissolving salt of NaCl and CaCl2 in 1 L of DI water. For heavy metals, solutions of 100 and 1000ppm concentrations were prepared by dissolving salt of Pb2+, Zn2+, Cu2+ (powdered with purity grade greater than 95%) in 1 L of DI water as it has been observed from the literature that the maximum heavy-metal contaminants concentration in a leachate is 1000 ppm (Prudent et al. 1996).

2.3 Determination of swelling behaviour The bentonites were mixed at their OMC and kept in humidity controlled desiccators for 24 hours to attain the moisture equilibrium. A conventional oedometer apparatus was used for determination of the

32

swelling behaviour of samples. The specimens were then statically compacted at their MDD in the consolidation ring of 60mm internal diameter and 20mm height. The specimens were positioned in the loading frame with a seating load of 4.9kPa. The soil samples were then inundated with water and the various salt solutions and allowed to swell till the swelling increment reached negligible values.

Table-1 Properties of bentonites used in this study

Property Bentonite-A Bentonite-B

Liquid limit 218.0% 560.0% Plastic limit 35.5% 36.0% Plasticity index 182.5 524.0 Shrinkage limit 16.3% 19.7% Specific gravity 2.8 2.82 Clay content 57.0% 68.0% Silt content 43.0% 32.0% Cation exchange capacity (CEC) (meq/100gm) 27.2 44.6 Exchangeable sodium percentage (ESP) 38.8% 54.2% Specific surface area (m2/g) 339.2 456.1 Optimum moisture content (OMC) 33.1% 32.1% pH 8.8 9.7 Maximum dry density (MDD) g/cc 1.23 1.28

Results and Discussions

3.1. Time swelling relationship in presence of inorganic salts Figures 1 and 2 show the relationship between the swelling of bentonite expressed in percentage and elapsed time for the samples in the presence of various concentrations of NaCl and CaCl2 solution. Irrespective of the type of saturating fluid, the time-swelling curve followed a “S” shape in which initially the bentonite swelled slowly, then the swelling increased steeply and reached to an asymptotic value. The plots show that the sample swells in three stages and named as initial, primary and secondary swelling (Rao et al., 2006; Mishra et al., 2008). In the first stage of swelling, termed as initial swelling, a marginal increase in swelling was observed for all the samples irrespective of their initial compaction condition and type of saturating liquid. In the second stage of swelling, termed as primary swelling, a significant amount of swelling was observed for all the samples. In the final stage of swelling, termed as the secondary swelling, again a marginal swelling was observed for all the samples. Primary swelling develops relatively rapidly, as it is linked to the rate of matric suction dissipation. Since the secondary swelling is controlled by diffusion of salts it develops more slowly. A longer period is also needed to complete secondary swelling, perhaps because ionic diffusion is affected by adsorption–desorption reactions (Shackelford & Daniel, 1991). Figures 1 and 2 show that with the increase in the salt concentrations, the time taken for the initial, primary, and secondary swelling reduces. The figures show that the time taken for the completion of the initial swelling was reduced from 35 to 11 and 5minutes due to an increase in the concentration from 0 to 1N of NaCl and CaCl2 solution, respectively, for Bentonite-B. For the similar increase in the concentration, the time taken for the completion of the primary swelling was reduced from 11,000 to 270 and 150minutes for 1N of NaCl and CaCl2 solution, respectively. Similar trend was also observed for Bentonite-A, where the initial swelling was reduced from 30 to 10 and 3minutes and primary swelling reduced from 650 to 250 and 200minutes due to an increase in the concentration from 0 to 1N of NaCl and CaCl2 solution, respectively. Similarly, with the increase in the concentration of the

33

inorganic salt solution from 0 to 1N of NaCl and CaCl2 solution the primary swelling was decreased from 47.6 to 16 and 15 %, respectively, for Bentonite-B. However, the reduction in the initial and secondary swelling was marginal and in the range of 1–5 % due to the increase in the inorganic salt solution concentration. Similar trend was also observed for Bentonite-A.

A comparison between the two bentonites in Figures 1 to 2 showed that at the same time elapsed, the percentage of swelling was more for Bentonite-B, which possess a higher CEC, ESP and SSA value, in comparison to Bentonite-A.

Fig.1 Time–swelling plot at different concentrations of CaCl2 solution for Bentonite-A and –B

Fig. 2 Time–swelling plot at different concentrations of NaCl solution for Bentonite-A and -B

3.2. Time swelling relationship in presence of heavy metals Figures 3 and 4 show the relationship between swelling of the bentonites expressed in percentage and elapsed time for bentonite samples in the presence of various concentrations of Pb2+, Zn2+, and Cu2+

solutions. Irrespective of the type of saturating fluid, the time-swelling curve followed a “S” shape in which initially bentonite swelled slowly, then increased steeply and finally reached to an asymptotic value. Similar to the samples with NaCl and CaCl2 solutions, the plots show the samples with heavy metals ions also swelled in three distinct stages of swelling, i.e. initial, primary and secondary swelling

0

10

20

30

40

50

60

0.1 10 1000 100000

Swel

ling

(%)

Time (minutes)

water0.01N CaCl20.1N CaCl21N CaCl2

___ Bentonite A----- Bentonite B

0

10

20

30

40

50

60

0.01 1 100 10000

Swel

ling

(%)

Time (minutes)

water0.01N NaCl0.1N NaCl1N NaCl

___ Bentonite A----- Bentonite B

34

(Rao et al., 2006; Mishra et al., 2008). Figures 3 and 4 show that with the increase in the heavy metal ion concentrations, the time taken for the initial, primary, and secondary swelling reduces. The figures shows that the time taken for the completion of the initial swelling was reduced from 35 to 15, 13, and 10 minutes due to an increase in the concentration from 0 to 1000 ppm of Pb2+, Zn2+ and Cu2+solution, respectively, for Bentonite-B. For the similar increase in the concentration, the time taken for the completion of the primary swelling was reduced from 11000 to 8900, 4450 and 2500 minutes for Pb2+, Zn2+ and Cu2+ solution, respectively. Similar trend was also observed for Bentonite-A, where the initial swelling was reduced from 30 to 13, 10 and 9 minutes and primary swelling reduced from 650 to 310, 280 and 260 minutes due to an increase in the concentration from 0 to 1000 ppm of Pb2+, Zn2+ and Cu2+ solution, respectively.

Fig.3 Time–swelling plot for Bentonite-A and -B in the presence of 1000 ppm concentration of heavy metals

A comparison between the two bentonites showed that at the same elapsed time, the percentage of swelling was higher for Bentonite-B in comparison to Bentonite-A. Irrespective of the bentonite type, swelling was marginally higher in presence of Pb2+solution in comparison to Zn2+ and Cu2+ solution.

Fig.4 Time–swelling plot for Bentonite-A and -B in the presence of 100 ppm concentration of heavy metals

0

10

20

30

40

50

60

0.1 1 10 100 1000 10000 100000

Swel

ling

(%)

Time (minutes)

Water1000 ppm Pb2+1000 ppm Zn2+1000 ppm Cu2+

___ Bentonite A----- Bentonite B

0

10

20

30

40

50

60

0.01 1 100 10000

Swel

ling

(%)

Time (minutes)

Water100 ppm Pb2+100 ppm Zn2+100 ppm Cu2+

___ Bentonite A----- Bentonite B

35

4. Conclusions From the time swelling plots, it was observed that with the increase in inorganic salt and heavy metal ion concentrations, the time taken for the primary swelling of the bentonite samples reduce. A comparison between the two bentonites showed that at the same elapsed time, the percentage of swelling was higher for Bentonite-B (with a higher liquid limit, CEC and SSA) in comparison to Bentonite-A of lower quality in presence of the various salts. Irrespective of the bentonite quality, swelling was least in presence of high concentration solutions.

References Dakshanamurthy, V. (1978).A new method to predict swelling using a hyperbolic equation, Geotechnical

Engineering, 9, 29-38.

Daniel, D.E. (1984). Predicting hydraulic conductivity of clay liners,Journal of Geotechnical Engineering, ASCE, 110(4), 465-478.

Lu, H. J., He, W., Liao, Z. W. and Chen, W. (2013). The Swelling, Shrinkage and Cracking Properties of Compacted Clay, Electronic Journal of Geotechnical Engineering, 18.

Madsen, F. T. and Vonmoos, M. (1989).The swelling behaviour of clays, Applied Clay Science, 4(2), 143–156.

Mishra, A.K., Ohtsubo, M., Li, L. and Higashi, T. (2005).Effect of salt concentrations on the permeability and compressibility of soil-bentonite mixtures, Journal of the faculty of agriculture, Kyushu university, Fukuoka, Japan, 50(2), 837-849.

Mishra, A. K., Ohtsubo, M., Li, L. and Higashi, T. (2011).Controlling factors of the swelling of various bentonites and their correlations with the hydraulic conductivity of soil-bentonite mixtures, Applied Clay Science, 52, 78–84.

Mishra, A.K., Dhawan, S. and Rao, S.M. (2008).Analysis of swelling and shrinkage behaviour of compacted clays, Geotechnical and Geological Engineering, 26(3), 289-298.

Mitchell J. K. and Soga, K. (2005).Fundamentals of soil behavior, 3rd edition, Wiley, New York.

Norrish, K. (1954). The swelling of montmorillonites, Discussions of Faraday Society, 18, 120–134.

Norrish, K. and Quirk, J. (1954).Crystalline swelling of montmorillonite; use of electrolyte to control swelling, Nature, 173, 255–257.

Ohtsubo, M., Li, L.Y., Yamaoka, S. and Higashi, T. (2004). Leachibility of heavy metals and salt frombottom ash. 5th Geoenvironmental Engineering Symposium, Japanese Geotechnical Society, 169–174.

Olson, R. E. and Mesri, G. (1970).Mechanisms controlling the compressibilityof clay, Journal of the Soil Mechanics and FoundationsDivision, ASCE, 96(6), 1863–1878.

Prudent, P., Domeizel, M. and Massiani, C. (1996). Chemical sequential extraction as a decision-making tool: application to municipal solid waste and its individual constituents, Science of the Total Environment, 178, 55-61.

Rao, S. M., Thyagaraj, T. and Thomas, H. R. (2006). Swelling of compacted clay under osmotic gradients, Geotechnique, 56 (10), 707–713.

Shackelford, C. D. and Daniel, D. E. (1991).Diffusion in saturated soil. 1: Background, Journal of Geotechnical Engineering, ASCE, 117(3), 467– 484.

Sivapullaiah, P. V., Sridharan, A. and Stalin, V. K. (1996).Swelling behaviour of soil bentonite mixtures, Canadian Geotechnical Journal, 33(5), 808-814.

36

SUITABILITY OF SLOPE STABILIZATION TECHNIQUES :

AN ASSESSMENT BASED ON CASE STUDIES Oindrila Das1, Ansel Jose2, Satya Ranjan Samal3

ABSTRACT Slopes in soils and rocks are found in nature and in-made structures. For natural slopes landslide is the dangerous phenomenon of slope failure which leads disaster to economy and human lives. Landslides may occur to various reason, such as earthquake, liquefaction, removal of vegetation etc. Failure in artificial slopes which are constructed to create highways, levees, railway track, dam, residential areas etc may be caused by design errors which include errors in slope inclination, slope height and inability to estimate the accurate load and the soil resistance. This paper reviews various case studies on implementation of slope stabilization methods such as introducing soil reinforcement, increasing vegetation, constructing retaining wall and using geotextiles. In tropical countries where landslide occurs in monsoon, geotextile addition and vegetation growth are found to be effective in stabilizing the slope. In residential areas even good drainage can arrest slope failure by reducing the pore water pressure. In hilly areas, where excavation has changed the original slope angle, making it prone to failure, reinforcement and retaining structure are observed as good solutions. For road and railway embankment stability, geotextiles are opted as a suitable solution.

1. INTRODUCTION Slopes are two types, such as natural slope and man-made slope or embankment. In modern day civilization slopes are being used for various purposes. Natural slopes and embankment are generally used as residential areas, agricultural field and transportation routes. Failure of natural slopes (i.e. landslide) is caused by earthquake, liquefaction, removal of vegetation, prolonged rainfall and seepage etc. Most of the failure of man-made slope is caused due to design errors, poor soil material, excessive cutting of slopes etc. So when a slope fails it greatly affects the daily life and economy of the concerned area. That makes stabilizing a slope is a great challenge for geotechnical engineers. 1 M.Tech Student, KIIT University 2, 3 Assistant Professor, KIIT University

37

There are many analytical methods for slope stabilization such as limit equilibrium method and finite element method. Generally a slope is determined to be stable if the factor of safety of it is more than 1. But according to the use and importance of the slope the factor of safety differs.

2. SLOPE STABILIZATION TECHNIQUE Slope stabilization generally involves reduction of driving force, increase in resisting forces or both. Driving forces can be reduced by excavating the materials from the proper part of the unstable ground and drainage of water to reduce the hydrostatic forces. Resisting force can be increased by elimination of weak strata, building of retaining structure, provision of in situ reinforcement, chemical treatment, growing vegetation etc. Also slope flattening, benching of slope, introducing counter beam at toes, introducing shear keys, soil nailing and geotextile placement are well established techniques as well.

3. DISCUSSION ON MAJOR METHODS Though there are several methods in stabilizing slope, some of them are widely used due to the design and economic benefits. Their mechanism is as follows.

Geotextile Reinforcement: - Geotextiles are planner, permeable and porous objects made of either jute, coir or polypropylene, polyethylene, polyester etc. It is basically a strong thick cloth. The capacity of reinforced layer is taken as either the allowable pull out resistance behind the potential failure surface or as its allowable design strength; whichever is less.

Slope Flattening: - In this method, the slope is excavated to turn it into gentler slope from steeper slope by changing its angle. Beside of reducing the sum of driving forces this method also tends to force the failure surface to move deeper into the ground.

Drainage in slope: - In many cases, slope fails due to saturation and excessive pore water pressure in subsoil. If a good drainage system can be provided it can drain out the excess water and prevent the slope from getting saturated.

Retaining structure: - They are provided to arrest the downward movement of soil mass. Retaining wall is one of the most popular retaining structure. It is used when cut or fill is required but there is no sufficient space available for just the slope itself. The walls are generally made deep enough so that the critical slip surfaces pass around it with an adequate FOS.

Soil nailing: - Soil nailing is a method of in-situ reinforcement utilizing passive inclusion, which will be mobilized if movement occurs. It can be used to retain, excavations and stabilize slope by creating in-situ soil reinforced retaining structure.

38

Vegetation: - Vegetation enhances stability of slope by hydraulic and mechanical mechanism. The root network of the vegetation works as natural soil reinforcement. By the root water uptake mechanism this method makes the slope stable by reducing pore water pressure and permeability and increasing the soil shear strength.

4. REVIEW ON CASE STUDIES Sutejo and Gofer (2015) [1] have discussed the case of a slope failure in an industrial area in Johor Bahru, Malaysia. In this case the slope was subjected to excessive cutting to make room for construction of factories. The original height of slope was varying from 20 m to 18 m and the slope angle was 43°. The slope was made of very dense silty sand whose angle of internal friction was 31°. The slope was barren to vegetation. The factor of safety was 1.28 for the modified slope having slope angle 70°. Due to the excavation activity, the shear strength of the soil was decreased and the slope became prone to failure. The rectification work included building a retaining structure at the toe of the slope and application of shotcrete at the face of the slope (Fig.1). FOS of slope is increased to 2.08 after the treatment.

Figure 1: Slope face after stabilization (Sutejo and Gofer, 2015)

Jamaluddin et. al (2011) [2] have studied the landslides of Hulu Langat (2011) and Bukit Antarabangsa (2008) in Malaysia. As it is a tropic region, most of the landslide occur in monsoon seasons that are induced by heavy rainfall and 80% of them are caused due to design and construction errors. Construction activities near the toe of a slope causes failure as lateral resistance is removed. In this case study, the method of slope flattening was adopted. As it was already a residential area, the slope was bounded with grass after reducing the slope angle. To avoid the increase in excess pore water pressure during heavy rainfall, good drainage system was introduced along with surface drain and weep hole.

39

Mulia and Prasetyorini (2013) [3] aims on stabilizing the upper basin of Ambang river. The basin was formed due to volcanic eruption which made this land fertile yet erodible. The basin area is used for forestation, agricultural, residential and transportation purpose. The slope formed by river stream is affected by land use conversion. The increasing activities in the land enhanced the load and seepage of groundwater along the slope. The slip surface of the slope is assumed to be circular and thickness of soil is more the 90cm. The adoption of slope stabilization method was varying depending on the land use. For forest are, reforestation was best option as it can arrest surface runoff and increase the water filtration to refill the ground water. For the agricultural area, improving drainage system could keep the slope stable by regulating pore water pressure. For residential areas, installing retaining structure combined with good drainage system was opted. Gabion meshes are used to stabilize slope in rural areas while sheet piling and geosynthetic reinforced wall are used to stabilize slope in narrow urban areas.

Mehrotra and Bhandari (1988) [4] have discussed about the landslide in Kaliasaur in Garhawal Himalaya located at the left bank of river Alakananda. According to the records of Geological Survey of India there are many cases of landslide in this particular area. Kaliasaur landslide is essentially multi-tiered, retrogressive landslide of complex rock formation with fault plane carrying the evidence of intense tectonic activities in past. Road construction activity and repeated cutting in slope for area development, poor drainage and recurring debris slide in the colluvium cover on the slope, removal of vegetation and erosion due to river action, all have been a great cause for the landslide. To solve this problem there were many methods adopted such as grading of slope, timber piling into the crack, construction of anchored diaphragm retaining wall and a toe wall at the junction of downhill and river. But most importantly the slope was re-vegetated to stabilize. After stitching the slope with timber pile, vegetation was established resulting dense root network penetrating about 1 meter in the slope anchor.

Chen et al. (2016) [5] have focused on multiple case in China and one among them is about Jeitai Temple which is 1400 years old, situated in a mountain. The ridge of the mountain is 1200 m long from south to north and averagely 450 m wide from west to east. As a result of southward mining of two coal dykes at a depth of 122m and 175 m an opening appeared beneath the fourth level of terrain. In addition to that due to heavy rainfall, penetrating fractured zones and cracks were observed at the temple yard causing slope deformation. The main measure of the stabilization method was to introduce multi-point anchored slope stabilizing pile of maximum length of 64m. Four locations on the side and middle of the landslide were chosen to set the reinforcing piles. 109 anchors were arranged with a total drilling depth of 5942.4 m. The graph below shows the reduction in landslide over years after treatment (Fig. 2).

of thanalyand reinf3000

Lancis mapropof slowall canti

obseraccumwas 840mmain1.55m3.5mdiamcemeA tot

Fi

Next casehe dam and ysis showedpre-stressed

forced with0kN re-stres

In the slocang River, ade of alluverties and rope, piles wwas forme

ilever stabil

During therved. The mulation boaxially 530

m. Accordinn sliding bem at the tai

mm in thicknmeter of 10ment ratio of tal of 18 ext

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d FOS of 0.d anchor ca

h 10 numbessed anchors

ope of Xiaoa remarkab

vial soil. Thrainfall. Depwere built ated which coizing pile. T

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displacements

h the Hongjor the proje8 i.e, unstabables were ers of 20ms.

owan hydroble slope fahe deformatpth of slidint the elevationtained 10The length o

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f the Yangtzdrilling holefrom 0.19mhe soil anchonical contoalled in the routed into ts were insta

40

s of Jietai Tem

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0 numbers oof the piles

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power projeigh beddingSuper large-d to support stabilizing

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nd its heigheen observeat the fron6m in leng

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(Chen, 201

ect. Due to g slope was-scale slopet the slope.g piles with

d of middleDecember cavation, blabout 25m. e slope stabi

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the requires formed ane stabilizing. The slopeh 270 piec

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4MPa. pe.

41

5. CONCLUSION In this recent days of overpopulation, mankind needs more space on the Earth and hence usage of slope as well as its stabilization have become important. Out of the several slope stabilization techniques few methods are discussed in this review. From the above case studies it has been observed that repeated cutting and filling of slope can make a naturally stable slope unstable. So it is important to design a slope or it's stabilization method with care. It is observed that in some cases application of shotcrete at toe rather than making a large retaining structure is more economical. For hilly areas use of slope stabilization piles are important as they help in reinforcing the soil. But along with, recreation of vegetation is necessary as it is both eco-friendly and economic solution. In tropical regions during monsoon excessive raining increases pore water pressure causing the slope to fail and hence designing a good drainage through the slope is found to be an effective solution.

6. Reference:- [1] Yulindasari, S. and Nurly, G. (2015), “Effect of Area Development on the

Stability of Cut Slopes”, Procedia engineering (Volume 125) Pages 331-337

[2] Mizal-Azzmi, N., Mohd-Noor, N. and Jamaludin, N. (2011), “Geotechnical Approaches for Slope Stabilization in Residential Area”, Procedia Engineering (Volume 20) Pages 474-482

[3] Adhi, Y., Mulia and Lucky, A. P. (2013), “Slope stabilization based on land use methods in Ambang sub river basin”, Procedia Environmental Sciences (Volume 17), Pages 240-247

[4] Mehrotra, G.S. and Bhandari, R. K. (1988), “A Geological Appraisal of Slope Instability and Proposed Remedial Measures at Kaliasaur Slide on National Highway, Garhwal Himalaya”, International Conference on Case Histories in Geotechnical Engineering

[5] Zuyu, C., Zhong, Z. and Chuangbing Z. (2016), “Recent advances in high slope reinforcement in China: Case studies”, Journal of Rock Mechanics and Geotechnical Engineering (Volume 8) Pages 775-788

[6] Jose, A., Patel, S. K., & Singh, B.(2015), "laboratory investigation on strength characteristics of fibre reinforced cohesive soil", Indian Geotechnical Conference, 2015, Pune, India,

42

EFFECT OF CONFINING PRESSURE AND VOID RATIOS ON SHEAR WAVE VELOCITY

FOR COHESIONLESS SOIL Dr. Rana Chattaraj, Mr. Prabhakar Kumar

and Ms. Neha Nasreen Introduction: Proper estimation of dynamic shear modulus is very much essential for the assessment of any soil structure interaction problem. Dynamic shear modulus of soil is generally found out in laboratory from the resonant column test. Shear wave velocity is obtained from the resonant column test and the dynamic shear modulus is calculated from the shear wave velocity. Thus the accurate estimation of shear wave velocity is at most priori to obtain the dynamic shear modulus of the soil. Shear wave velocity for the cohesionless soil mostly depends on confining pressure and void ratio. In this paper an effort has been made to study the effect of confining pressure and void ratio on dynamic shear modulus. Based on the laboratory test, an empirical formula has been suggested to estimate the shear wave velocity. This empirical formula may be utilized for crude estimation of shear wave velocity for cohesionless soil where proper laboratory or field test data is not available.

Keywords: Dynamic shear modulus; Shear wave velocity; Resonant column test; Empirical relation.

Literature review: Various researchers around the world have initiated laboratory test program as well as field test program to estimate the dynamic properties of the soil. Hardin (1978) has given a formula for estimation of Gmax for sand and clay. Chung, et al. (1986) has conducted resonant column test on Monterey sand 0 to study the dynamic properties of the sand at small strain. Saxena and Reddy (1989) have also performed the resonant column tests on Monterey sand 0. . De Alba, et al. (1984), Tokimatsu and Uchida (1990) have given correlation between the liquefaction potential of a soil and the shear wave velocity. Though numerous researchers have studied the dynamic properties of soil and suggested empirical relations, still there is a scope to study the dynamic properties of soil to strengthen the literature as the properties of soil differ from place to place.

Soil properties: For this experimental investigation, Kasai River sand is taken. Based on the laboratory investigation, the following index properties of the Kasai River sand are obtained. Specific Gravity = 2.64, Coarse sand = 1.7%, Medium sand = 44%, Fine sand = 54% and Fine content = 0.3%. Maximim void ratio = 0.83, Minimum void ratio = 0.56, Coefficient of Uniformity = 2.36, Coefficient of Curvature = 1.08. According to USCS, this sand is classified as poorly graded sand (SP).

43

Sample preparation: A split mould with 70mm internal diameter and 140mm height is used to prepare the test samples. All the samples are prepared by tamping method as per ASTM D 5311-11 guidelines in three layers with oven dried soil.

Test procedure: A fixed-free type resonant column device was used for the test program. The detail test procedure can be found in anywhere else (Chattaraj. R and Sengupta. A- 2016). Resonant column test were conducted at four different confining pressure, such as 50, 100, 200 and 400 kPa and at four different void ratio of 0.61, 0.67, 0.71 and 0.76.

Result and discussion: Figure-1 represents the variation of shear wave velocity with strain at different confining pressure for the soil with void ratio (e) = 0.61. From the figure it is evident that the variation in shear wave velocity below 0.001% strain level is insignificant irrespective of the confining pressure. Thus shear wave velocity correspond to 0.001% strain is taken as the maximum shear wave velocity. Therefore, shear wave velocity corresponds to 0.001% of strain is used to calculate the maximum dynamic shear modulus (Gmax) for this respective soil. It is also evident from the graph that the shear wave velocity decreases nonlinearly with the increase in strain. Similar observation was also made for the soil with different void ratios. Thus it is concluded that at constant void ratio and confining pressure the shear wave velocity is a function of strain.

Figure-1 : Variation of shear wave velocity with strain at different confining pressure for e = 0.61

200

240

280

320

360

400

0.0001 0.001 0.01 0.1

Shea

r wav

e vel

ocity

(m

/s)

Strain (%)

Variation of shear wave velocity at various confining pressure

400 kPa

200 kPa

100 kPa

50 kPa

44

Effect of confining pressure on maximum shear wave velocity: Figure-2 shows the variation of maximum shear wave velocity (shear wave velocity corresponding to 0.001% strain) with confining pressure at constant void ratio. It can be seen from the figure that the increase in confining pressure results in increase in maximum shear wave velocity. It can be attributed to the fact that as the confining pressure increases, the soil becomes stiffer as the void ratio decreases. Thus, more and more contact points are generated and as well as contact area amongst the particles are increases. Increase stiffness with increase in contact points facilitate the shear wave to move faster. Thus increase in confining pressure eventually increases the shear wave velocity, which directly increases the maximum dynamic shear modulus. Change in confining pressure from 50 kPa to 400 kPa resulting in a 55.6% increase in shear wave velocity for the soil with void ratio e = 0.61.

Effect of void ratio on maximum shear wave velocity: Figure-3 represents the effect of void ratio on maximum shear wave velocity at constant confining pressure. It is evident from the graph that as the void ratio increases, the corresponding maximum shear modulus decreases. With the increase in void ratio, the soil become loose thus the stiffness of the soil decreases. Decrease in stiffness is the main reason behind the decrease in shear wave velocity. This kind of observation was also made for the soil subjected to different confining pressure.

Figure-2: Variation of shear wave velocity with confining pressure for e = 0.61

200

250

300

350

400

0 100 200 300 400 500

Shea

r wav

e vel

ocity

(m

/s)

Confining pressure (kPa)

Void Ratio = 0.61

45

Figure-3: Variation of shear wave velocity with void ratio for 400 kPa

Correlating maximum shear wave velocity, confining pressure and void ratio: Based on the result obtained from the resonant column test, an empirical relation was developed to predict the maximum shear wave velocity. The following relation is given below.

Maximum shear wave velocity = )7.03.0(

)()(18.1592

766.00

5.0

ePa+

×× σ

Where, Pa is the atmospheric pressure. The unit of Pa is same as that of the effective confining pressure, σ0, and e is the void ratio. As the above equation is dimensionally correct, this equation may be used in any system of units. The coefficient of determination (R2) is found to be 0.91 for this proposed relationship.

Figure-4: Comparison of predicted and observed values

of maximum shear wave velocity

330

335

340

345

350

355

0.6 0.65 0.7 0.75 0.8

Shea

r wav

e vel

ocity

(m

/s)

Void ratio(e)

Confining pressure = 400 kPa

100

200

300

400

100 200 300 400Prop

osed

val

ue (m

/s)

Observed value (m/s)

46

Validation of the proposed statistical model: Another set of test result which was not used for the statistical analysis was used to cheek the validity of the proposed equation. The other set of test result was compared with the values obtained from the proposed equation at similar condition. Less than 10% variation was observed from the values obtained from the proposed equation.

Conclusion: It is observed that the void ratio and effective confining pressure greatly influence the shear wave velocity of cohesionless soil. As it is seen that the proposed empirical equation closely approximate the test data, thus this empirical equation may be used for the prediction of shear wave velocity for Kasai River sand in the range of 50-400 kPa confining pressure with the void ratios ranges from 0.61 to 0.76 for all practical purposes. This empirical equation also may be used for the other type of sand for the crude estimation of maximum shear wave velocity in the proposed range.

References:

ASTM Standard D-5311-11: Standard Test Method for Load Controlled Cyclic Triaxial Strength of Soil. American Society of Testing and Materials, West Conshohocken, Pennsylvania, USA

Chattaraj, R., and Sengupta, A. (2016). “Liquefaction potential and strain dependent dynamic properties of Kasai River sand”. Soil Dynamics and Earthquake Engineering, Elsevier. 90, 467-475.

Chung, R. M., Yokel, F. Y., and Drnevich, V. P. (1984). "Evaluation of dynamic properties of sand by resonant column testing." Geotechnical Testing Journal, GTJODJ, 7(2), 60-69.

De Alba, P., Baldwin, K., Janoo, V., Roe, G., and Celikkol, B. (1984). "Elastic-wave velocities and liquefaction potential." Geotechnical Testing Journal, GTJODJ, 7(2), 77-87.

Hardin, B. O. (1978). "The nature of stress-strain behaviour for soils." Proc. of Earthquake Engineering and Soil Dynamics, ASCE, Pasadena, CA.1, 3-90.

Saxena, S. K., and Reddy, K. R. (1989). "Dynamic moduli and damping ratio for Monterey No.0 sand by Resonant Column tests." Soils and Foundation, 29 (2), 37-51.

Tokimatsu, K., and Uchida, A. (1990). "Correlation between liquefaction resistance and shear wave velocity." Soils and Foundation, 30(2), 33-42.

47

STRENGTH PARAMETER UNDER

PLANE STRAIN CONDITION

Dr. Satyajeet Nanda, Dr. Benu Gopal Mahapatra and Mrs. Bandita Paikaray

Abstract

The strength parameters (C & φ ) significantly varies with mode of shearing. It has been established by various studies that plane strain condition produces higher φ

value compare to that of triaxial condition. Many geotechnical structures experience plane strain condition. Since availability of plane strain test set up is rarer, designers usually use strength parameters obtained under triaxial condition. This paper preset a mathematical method to predict strength parameters at plane strain condition once strength parameter under triaxial condition is known.

Introduction : The geotechnical professional often use triaxial test to determine the soil properties. Many of geotechnical structures are exhibiting plane strain condition. The soil properties vary significantly between triaxial and plane strain condition. Figure 1 (a) and Figure 1 (b) show typical stress-strain curve for plane strain and triaxial condition. Lee (1970) observed the difference in φp between plane strain and triaxial up to 8o and the triaxial condition produces more than twice of γp compared to the plane strain condition.

(a)

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(b) Figure 1. Stress strain relationship for plane strain and triaxial specimens (a) Monterey sand (adopted from Marachi et al. 1981) (b) Antioch sand (adopted from Lee 1970)

Dense sand shows higher change in shear strength parameters compared to the loose soil. Lee (1970) and Marachi et al. (1981) also observed that at higher confining pressure the difference between plane strain and triaxial condition decreases. In general, the plane strain apparatus is complex compare to the triaxial and not available in majority of soil laboratories. In this paper, a method is proposed to predict angle of friction for plane strain condition from triaxial friction angle. In this method it is assumed that the maximum angle dilatancy remains same for triaxial and plane strain condition.

The variation in φ between plane strain and triaxial condition Several attempts have been made to relate shear strength parameters (φc= internal friction angle at critical state; φp= internal friction angle peak critical) with the dilatancy angle. Rowe (1962) developed stress-dilatancy theory based on hypothesis that energy losses during shearing and minimum energy ratio will be achieved at the failure. He proposed shear strength and dilatancy relationship can be expressed as

tan 45 tan 45 tan 45 (1)

Where, is the Rowe’s friction angle.

Schanz and Vermeer (1996) proposed that the dilatancy angle can be defined in a single equation for plane strain and triaxial condition:

sin (2)

Soil is a granular material hence the dilation of soil mass depends on the density and arrangement of particle structure. The process of dilation becomes more complicated as experimental observations indicate particle crushing at higher confining

49

pressure. After analyzing large number of experimental data Bolton (1986) proposed a relationship between dilation, soil density and stress parameter as

υ AI (3)

Where A= constant; I = Dilatancy index. Further Bolton (1986) proposed an empirical relationship for I

I ID Q ln ′ R (4)

ln p Q RID (5)

Where ID= relative density; p = stress just sufficient to eliminate dilation by crushing; p′= mean effective stress at failure; p = reference stress; Q and R are the fitting parameters. Using unit of stress as kPa and analyzing large amount of laboratory data, Bolton (1986) observed that the values of Q and R are 10 and 1 respectively and are found to be best fit with the various soils. Salgado et al. (2000) reported the value of Q and R may vary with silt content in the range of 7.3 to 11.4 and 1.29 to 0.49 for Q and R respectively. The relation between peak and critical friction angle as a function of relative density as given by Bolton (1986) can be written as

AψI (6)

A is 5 for plane strain and 3 for triaxial condition. Further, Bolton (1986) given the following expression for plane strain condition

P P 5I 0.8υ (7)

and for triaxial condition

T T 3I 0.5υ (8)

The superscript P and T represent plane strain and triaxial condition.

Assuming the maximum angle of dilation remains same for plane strain and triaxial condition and from equation (7) and equation (8) the relation between peak and critical friction angle can be expressed as given in equation 9 and 10.

P P APυ (9)

And

T T ATυ (10)

The ratio between critical and peak friction angle for plain strain and triaxial can be expressed in terms of KφP and KφT respectively as given in equation 11 and 12 respectively.

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K P PP (11)

K T TT (12)

Combining equation (9) to (12)

P T APAT K TK P (13)

Equation (13) can be used to calculate P . K T and T can be determined from

regular triaxial test. The APAT and K P are unknown. Number of published literature has

been studied to established the relationships between K T, APAT and K P. Figure 2 to 4

show the above relationships and Table 1 shows the details of soils data used. Figure 2 shows the relationship between K T and K P. The value of K T is always higher than or equal to the K P. The observed variation in Figure 2 can be expressed empirically as

K P 0.19e . K T (14)

Figure 3 shows the variation in K T with nr. nr is the relative porosity which can be defined as

n (15)

Where, nmax and nmin are the maximum and minimum porosity of the sand respectively. nr varies from 1 to 0 for dense to loose state of sand. From Figure 3 it is seen that the K T decreased with the increase in soil density. Higher confining pressure

may increase the K T. The variation in APATwith K T is shown in Figure 4. The value of APAT varies from 1.5 to 1.3 for dense to loose soil at higher confining pressure. For

medium confining pressure the value of APAT varies with very narrow margin of about

1.55 to 1.45 for dense to loose soil. At lower confining pressure APAT is almost constant to

1.6. Bolton (1986) reported the value of APAT is about 1.6. From the laboratory

observations reported in literature, it has been found that for a reasonable prediction,

the values of APAT can be taken in the range of 1.5 to 1.6 for K T greater than 0.7, and

1.4 to 1.5 for K T less than 0.7.

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Figure 2 . Relationship between K T and K P

Table 3.1. The source of soils data used for formulation

Identification Confining pressure

emax emin G D50 Cu Reference

A 344 0.8 0.4 2.65 0.22 2.4 Hanna (2001)

B 344 0.9 0.5 2.63 0.65 2.33 Hanna (2001)

C 344 0.95 0.4 2.64 0.65 2 Hanna (2001)

A1 172 0.8 0.4 2.65 0.22 2.4 Hanna (2001)

B1 172 0.9 0.5 2.63 0.65 2.33 Hanna (2001)

C1 172 0.95 0.4 2.64 0.65 2 Hanna (2001)

BR 276 0.79 0.47 2.68 0.25 2.42 Cornforth (1964)

M 28 0.82 0.49 2.655 0.2 1.8 Rowe (1969)

G is the specific gravity; Cu is the coefficient of uniformity

The P and P can be determined from triaxial test result by using equation (13), equation (14) and Figure 2 to 4. The value of K T should be know from the triaxial test.

The value of K P can be obtained from equation (14). The value of APAT can be fixed from

Figure 4 which is ranging 1.4 to 1.6 at various conditions as discussed above. Once the

value APAT and K P are known, the P can be determined from equation (13) and P

from equation (11).

0.5

0.6

0.7

0.8

0.9

1

1.1

0.7 0.75 0.8 0.85 0.9 0.95 1 1.05

Kφ

P

KφT

ABCA1B1C1BRM

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Figure 3. Variation in K T with nr

Figure 4. Variation in APAT with K T

0.7

0.75

0.8

0.85

0.9

0.95

1

0 0.2 0.4 0.6 0.8 1 1.2

KΦ

T

nr

A

B

C

A1

B1

0.7

0.75

0.8

0.85

0.9

0.95

1

1.2 1.3 1.4 1.5 1.6 1.7 1.8

KΦ

T

Ap/AT

ABCA1B1C1BRM

53

Conclusion

A mathematical procedure has been proposed to predict angle of friction for plane strain condition. For this prediction require input parameters is angle of friction and dilatancy angle, which must be obtained under triaxial condition. The proposed formulation was designed by taking observations from various laboratory investigation reported in literature. Two new parameters K T and K P was defined and interrelationship between theses parameters have been derived. Various charts (Figures 3 & 4) and interrelationship (K T and K P) can be used to predict angle of internal friction angle for plane strain condition.

References

Bolton, M.D. (1986). “The strength and dilatancy of sands.” Geotechnique, 36(1), 65 – 78

Lee, K.L. (1970). “Comparison of plane strain and triaxial tests on sand.” J. Soil Mech. Found. Div., 96(SM3), 901 – 923.

Marachi, N.D., Duncan, J. M. and Seed, H.B. (1981). “Plane-strain testing of sand.” Laboratory Shear Strength of Soil. ASTM STP 740. R. N.Yong and F. C. Townsend, Eds., American Society for Testing and Materials, 294-302.

Rowe, P.W. (1962). “The stress-dilatancy relation for static equilibrium of an assembly of particles in contact.” Proc. R. Soc. London, A 1962 269, 500-527.

Salgado, R., Bandini, P. and Karim, A. (2000). “Shear strength and stiffness of silty sand.” J of Geotech. Geoenviron. Eng., 126(5), 451-462.

Schanz, T. and Vermeer, P. A. (1996). “Angles of friction and dilatancy of sand.” Geotechnique 46(1), 145-151.

54

EMPERICAL STUDY OF AXIALY LOADED ROCK SOCKETED PILE IN BHUBANESWAR REGION

Hemanta K. Dash, Aradhana Mishra and Sitaram Satapathy

ABSTRACT: This paper presents the analysis of static load tests that is carried out on axially loaded piles in Bhubaneswar region of Odisha. The load settlement behavior for different diameter piles is plotted and ultimate pile capacity is estimated by using different empirical methods. The safe load is calculated by using the criteria given in IS-14593. The variation of ultimate load w.r.t. pile diameter for Bhubaneswar region can be used to estimate the ultimate pile capacity of large diameter pile which cannot be tested up to failure.

Keywords: Pile Load Test, Load Settlement Curves, Empirical Methods, Ultimate load, Rock Socketed Piles

INTRODUCTION : The use of drilled piles socketed into rock as foundation structures is one of the best solutions when layers of loose soil overlie bedrock at shallow depths. In these cases, considerable bearing capacity can be ensured by the shaft friction in rock, even with small pile displacements (Carrubba, 1997). The axial load carrying capacity of rock-socketed cast-in place piles can be estimated by applying static analyses, information/data collected from pile load tests, numerical methods and empirical approaches. Load tests are conducted to determine the in situ bearing capacity and the load– deformation behavior of piles. Pile load testing provides the most reliable information for the design, because it is a large scale, if not full-scale, model for the behavior of a designed pile in actual soil conditions.

PILE LOAD TEST : Pile load test is to determine the settlement under working load and determine ultimate capacity of pile on ground.In general two types of pile load tests are conducted. They are:

o Initial Test o Routine Test

The initial test is performed before the start of construction to assess the design adequacy. The routine test is performed on a working pile. This test is also known as work test. In initial test, the test load 5/2 times the working load in work test load is 3/2 times the design load.

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The pile load test data obtained from pile load test conducted at kamakhysa nagar of Bhubaneswar is collected from S M consultancy, Mancheswar, Bhubaneswar.

CRITERION FOR SAFE LOAD (IS 14593 - 1998) : Rock socketed piles are designed to carry compressive loads either in side shear or end bearing or combination of the both. This criterion is recommended for computing safe load capacity of the rock socketed pile. Safe load capacity is also evaluated from the field load settlement data. The load settlement curve is extrapolated for the regression value R2 close to 1.

As per IS 14593 (1998), the safe load for a socketed pile is considered as the minimum of the following:

1. Fifty percentage of the load at 12 mm settlement. 2. One third of the ultimate failure load.

METHODS TO DETERMINE ULTIMATE LOADS :As per above criterion the safe load also depends on the 1/3 of the ultimate load. To evaluate the ultimate load various empirical methods have been proposed in the literature to determine the bearing capacity of piles using the results of pile load tests

1. DE BEER YIELD METHOD (1968) : De Beer made use of the logarithmic linearity by plotting the load-movement data in a double-logarithmic diagram. The intersection point of two straight lines on a log-log plot gives the magnitude of ultimate load. But the use of this method has a constraint due to the reason that, in most of the load tests, pile is not loaded upto failure.

2.CHIN’S METHOD (1970): Chin’s proposed to divide each movement with its corresponding load and plot the resulting value against the movement. After some initial variation, the plotted values will fall on straight line .The inverse slope of this line is the Chin-Kondner Extrapolation of the ultimate load.

3. MAZURKIEWICZ METHOD : Mazurkiewicz(1972) proposed a method that allows the failure load tobe extrapolated, even if the maximum test load is smaller than the failure load. Fig.2b shows how a set of equal settlement lines are chosen and the corresponding load lines are constructed. From the intersection of each load line with the load axis, a 45 degree line is drawn to intersect the next load line. This intersection fall approximately on a straight line which intersects the load axis at failure load. This method is based on the

56

assumption that the load-settlement curve is parabolic after an initial straight portion (parabolic criterion)

4.TANGENT-TANGENT METHOD : Applying tangent—tangent method, a plot is made between load divided by cross sectional area of pile and the settlement on semi logarithmic scale. The method plots two tangential lines along the initial and latter portions of the load – displacement curve.The intersection of two straight line is referred to as the failure load.

5.HANSEN METHOD : Applying Hansen Method the square root of each settlement value from field load test data divided by the cor- responding load value is plotted against the settlement. Estimation of the ultimate load by Hansen Method is given by the formula :

Qu =(2C1*C2)^1/2

Where Qu = ultimate load capacity C1 = slope of the best fitting straight line. C2 = y-intercept of the straight line

CONCLUSION : From the analysis of pile load test data using different empirical method it is found that all method except chin’s method gives satisfactory result. The Mazurkiewicz Method provided the maximum value for the ultimate load capacity. De Beer method and Tangent-Tangent Method gives suitable result for Bhubaneswar region.

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IMPORTANCE OF NABL ACCREDITATION AND PROCEDURE

Venugopal C1 & Er. Laxmi Kanta Tripathy2

Abstract Geotechnical investigation and soil testing is being carried out by all the laboratories which consists of qualified trained scientific personals, technicians and calibrated instruments. The instruments are being used for its purpose are also needs to be calibrated from time to time. The standard operating procedures (SOP) for testing of materials are also having its own importance for conducting the testing works. There are numerous BIS, ASTM, British & Euro codes etc. available and are to be followed for each of the testing purpose of soil and other materials. From time to time the codes are being reaffirmed and updated for the purpose depending upon new innovations. But due to some ignorance, sometimes the practices of updating by the laboratories are being ignored. NABL is the National Accreditation Board for Laboratories, which gives the right direction and standards for conducting the tests. May the laboratory belong to Government or Private, if the Man, Material, Method or Machineries are not correct, the test report produced by the laboratory is may not be correct one. So, NABL plays a major role for qualifying the laboratory up to the standard. In this article we can find the role played by the NABL & its procedure.

Introduction NABL is a constituent Board of Quality Council of India (QCI). QCI is a registered society under the Societies Registration Act, 1860. Department of Industrial Policy and Promotion, Ministry of Commerce and Industry, Government of India is the nodal Department for QCI.

NABL has been established with the objective of providing Government, Industry Associations and Industry in general with a scheme of Conformity Assessment Body’s accreditation which Involves third-party assessment of the technical competence of testing including medical and calibration laboratories, proficiency testing providers and reference material producers.

The laboratory accreditation services to testing and calibration laboratories are provided in accordance with:

1 Joint Director, NABL-QCI 2 Hony.Secy, IGS. Bhubaneswar

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• ISO/ IEC 17025: 2005 ‘General Requirements for the Competence of Testing and Calibration Laboratories’

• ISO 15189: 2012 ‘Medical laboratories -- Requirements for quality and competence’

• ISO/IEC 17043 :2010 “Conformity assessment -- General requirements for proficiency testing”

• Reference Material Producers based on ISO Guide 34:2009 / ISO 17034:2016 - General requirements for the competence of reference material producers

NABL offers accreditation services in a non-discriminatory manner. These services are accessible to all testing including medical and calibration laboratories, proficiency testing providers and reference material producers in India and other countries in the region, regardless of the size of the applicant CAB or its membership of any association or group or number of CABs already accredited by NABL.

NABL has established its accreditation system in accordance with ISO/ IEC 17011:2004 ‘Conformity Assessment – General requirements for Accreditation bodies accrediting conformity assessment bodies’.

NABL is an MRA signatory to Asia Pacific Laboratory Accreditation Cooperation (APLAC) and International Laboratory Accreditation Cooperation (ILAC).

The accreditation services are offered for construction industry includes building materials, Mechanical Properties of Metals, Soil/Rock and NDT.

Why Accreditation? Accreditation is the third party attestation related to a conformity assessment body conveying the formal demonstration of its competence to carry out specific conformity assessment task. Conformity Assessment Body (CAB) is a body which includes Testing including medical Laboratory, Calibration Laboratory, Proficiency Testing Provider, and Certified Reference Material Producer.

The liberalization of trade and industry policies of the Government of India has created quality consciousness in domestic trade and provided greater thrust for export. As a consequence testing centres and laboratories have to demonstrably operate at an internationally acceptable level of competence.

Laboratory accreditation is a procedure by which an authoritative body gives formal recognition of technical competence for specific tests/ measurements, based on third party assessment and following international standards.

Similarly, Proficiency testing Provider accreditation gives formal recognition of competence for organizations that provide proficiency testing. Reference Material

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Producers Accreditation gives formal recognition of competence to carry out the production of reference materials based on third party assessment and following international standards.

Benefits of Accreditation Formal recognition of competence of a laboratory by an Accreditation body in accordance with international criteria has many advantages:

1. Increased confidence in Testing/ Calibration Reports issued by the laboratory 2. Better control of laboratory operations and feedback to laboratories as to

whether they have sound Quality Assurance System and are technically competent

3. Potential increase in business due to enhanced customer confidence and satisfaction.

4. Customers can search and identify the laboratories accredited by NABL for their specific requirements from the NABL Web-site or Directory of Accredited Laboratories

5. Users of accredited laboratories enjoy greater access for their products, in both domestic and international markets.

6. Savings in terms of time and money due to reduction or elimination of the need for re-testing of products.

Procedure

• The CAB is required to apply in the prescribed application form (NABL 151 for testing laboratories, NABL 152 for calibration laboratories, NABL 153 for medical laboratories, NABL 180 for PTP and NABL 190 for RMP), in three copies along with two copies of the quality manual of the CAB that should describe the management system in accordance with ISO/ IEC 17025: 2005 or ISO 15189: 2012 or ISO/IEC 17043:2010 or ISO Guide 34:2009 whichever is applicable. The application is to be accompanied with the prescribed application fee as detailed in NABL 100. CAB has to take special care in filling the scope of accreditation for which the CAB wishes to apply. In case, the CAB finds any clause (in part or full) not applicable to the CAB, it is expected to furnish the reasons.

• NABL Secretariat on receipt of application form, the quality manual and the fees issues an acknowledgement to the CAB indicating a unique ID number, which is used for correspondence with the CAB. After scrutiny of application for its completeness in all respects, NABL Secretariat may ask for additional information/ clarification(s), if necessary.

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• In case there are no inadequacies in the quality manual or after satisfactory corrective action by the CAB, a pre -assessment visit of the CAB is organised by lead assessor appointed by NABL. The pre-assessment of the CAB is conducted to evaluate non-conformities (if any) in the implementation of the quality system, to assess the degree of preparedness of the CAB for the assessment, to determine the number of assessors required in various fields based on the scope of accreditation, number of key location to be visited etc. The lead assessor submits a pre-assessment report to NABL Secretariat with a copy to the CAB. The CAB takes corrective actions on the non-conformities raised on the documented management system and its implementation and submits a report to NABL Secretariat.

• After the CAB has taken satisfactory corrective actions, NABL finalizes the constitution of assessment team in consultation with the CAB. The team includes the lead assessor and technical assessor(s)/ expert(s) in order to cover various fields/ disciplines/ groups within the scope of accreditation sought. NABL may also nominate an observer. The assessment team reviews the CAB’s documented management system and verifies its compliance with the requirements of ISO/ IEC 17025: 2005 or ISO 15189: 2012 or ISO/IEC 17043:2010 or ISO Guide 34:2009 whichever is applicable and relevant specific criteria and other NABL policies. The CAB’s technical competence to perform specific tasks is also evaluated. The non-conformities if identified are reported in the assessment report. It also provides a recommendation towards grant of accreditation or otherwise. The report prepared by the assessment team is sent to NABL Secretariat. However a copy of summary of assessment report and copies of non-conformities if any, are provided to the CAB at the end of the assessment visit.

• The assessment report is examined by NABL Secretariat and follow up action as required is initiated. CAB has to take necessary corrective action on non-conformities and submit a report to NABL Secretariat within 60 days. NABL monitors the progress of closing of non-conformities.

• After satisfactory corrective action by the CAB, the Accreditation Committee examines the assessment report, additional information received from the CAB and the consequent verification, if any. In case everything is in order, the Accreditation Committee makes appropriate recommendations regarding accreditation of the CAB to the Chairman, NABL.

• All decision taken by NABL are open to appeal by the CAB. The appeal is to be addressed to the Director, NABL.

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• When the recommendation results in the grant of accreditation, NABL issues an accreditation certificate which has an unique number and NABL hologram, discipline, date of validity along with the scope of accreditation.

• For site laboratory, tests/ calibrations performed at site are clearly identified in the scope of accreditation while issuing the certificate.

• The applicant CAB must make all payments due to NABL, before the accreditation certificate(s) is/ are issued to them.

• The accredited CABs at all times shall conform to the requirements of ISO/ IEC 17025: 2005 or ISO 15189: 2012 or ISO/IEC 17043:2010 or ISO Guide 34:2009 whichever is applicable and relevant specific criteria and NABL Policies. The accredited CABs are required to comply at all times with the terms and conditions of NABL given in NABL 131 "Terms & Conditions for obtaining and maintaining NABL Accreditation".

• The NABL accreditation certificate is valid for a period of 2 years. NABL conducts annual Surveillance of the CAB at intervals of one year which is aimed at evaluating continued compliance to the requirements of ISO/ IEC 17025: 2005 or ISO 15189: 2012 or ISO/IEC 17043:2010 or ISO Guide 34:2009 whichever is applicable and relevant specific criteria and NABL Policies.

• The accredited CAB is subjected to re-assessment every 2 years. The CAB has to apply 6 months before the expiry of accreditation to allow NABL to organise assessment of the CAB, so that the continuity of the accreditation status is maintained.

Conclusion In the present scenario, Indian Geotechnical Society is being associated with NABL for updating the standards of Geotechnical testing. To get the test result in accurate manner and up to the world standard, we should encourage the Govt & Private Laboratories for NABL accreditation. As the good standard of testing has its own importance and this influences the standard and durability of infrastructures, it is the need of the time that, we should be going for it. As the reference of a quality control or quality assured product should be from a qualitative laboratory. NABL plays a major role for guiding the laboratory towards a right direction.

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CODES AND STANDARDS RELATED TO DEEP FOUNDATIONS

Indian Standards Codes

IS 2911: Part 1: Sec 1: 2010 Code of practice for design and construction of pile foundations: Part 1 Concrete piles, Section 1 Driven cast in-situ concrete piles

IS 2911: Part 1: Sec 2: 2010 Code of practice for design and construction of pile foundations: Part 1 Concrete piles, Section 2 Bored cast-in-situ piles

IS 2911: Part 1: Sec 3: 2010 Code of practice for design and construction of pile foundations: Part 1 Concrete piles, Section 3 Driven precast concrete piles

IS 2911: Part 1: Sec 4: 2010 Code of practice for design and construction of pile foundations: Part 1 concrete piles, Section 4 Bored precast concrete piles

IS 2911: Part 2: 1980 Code of practice for design and construction of pile foundations: Part 2 Timber piles

IS 2911: Part 3: 1980 Code of practice for design and construction of pile foundations: Part 3 Under reamed piles

IS 2911: Part 4: 1985 Code of practice for design and construction of pile foundations: Part 4 Load test on piles

IS 3267: 1981 General Requirements for anchors

IS 4999: 1991 Recommendation for grouting of pervious soils

IS 5121: 1969 Safety code for piling and other deep foundations

IS: 6186 – 1986 Specification for Bentonite

IS 6426: 1972 Specification for pile driving hammer

IS 6427: 1972 Glossary of Terms Relating to Pile Driving Equipment

IS 6428: 1972 Specification for pile frame

IS 6433: 1972 Specification for Guniting Equipment

IS 1892: 1979 Code of practice for sub-surface investigations for foundations

IS: 8009 (Part II)-1980 Code of Practice for Calculation of Settlement of Foundations Part II Deep Foundations Subjected to Symmetrical Static Vertical Loading

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IS: 9012 - 1978 Recommended Practice for Shotcreting

IS 9527: Part 3: 1983 Code of practice for design and construction of port and harbour structures: Part 3 Sheet pile walls

IS 9527: Part 4: 1980 Code of practice for design and construction of port and harbour structures: Part 4 Cellular sheet pile structures

IS: 9556- 1980 Code of practice for design and construction of Diaphragm wall

IS 9716: 1981 Guide for lateral dynamic load test on piles

IS 10270:1982 Guidelines for construction of Pre-stressed Rock Anchor.

IS 10492:1983 General requirements and testing of anchor cable stoppers

IS 11309:1985 Method for Conducting Pull-Out Test On Anchor Bars And Rock Bolts

IS 12584: 1989 Specification for bentonite for grouting in civil engineering works

IS 13094:1992 Selection Of Ground Improvement Techniques For Foundation In Weak Soils – Guidelines

IS 14343 : 1996 Guidelines for choice of grouting materials for alluvial grouting

IS 14362: 1996 Pile boring equipment - General Requirement

IS 14593: 1998 Design and construction of bored cast-in-situ piles founded on rocks – Guidelines

IS 14893: 2001 Non-Destructive Integrity Testing of Piles (NDT) – Guidelines

IS 15284 (Part 1): 2003 Design and Construction for Ground Improvement — Guidelines \

Part 1 Stone Columns

IS 15284 (Part 2): 2004 Design and Construction for Ground Improvement — Guidelines \

Part - 2 Preconsolidation using Vertical Drain

ASTM Codes:

D1143-81(2013) Standard Test Method for Piles Under Static Axial Compressive Load

D3689-90(2013) Standard Test Method for Individual Piles under Static Axial Tensile Load

D3966-90(2013) Standard Test Method for Piles under Lateral Loads

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D4380 - 12 Standard Test Method for Density of Bentonitic Slurries

D4435 - 08Standard Test Method for Rock Bolt Anchor Pull Test

D4436 - 08Standard Test Method for Rock Bolt Long-Term Load Retention Test

D4381 / D4381M - 12Standard Test Method for Sand Content by Volume of Bentonitic Slurries

D4945-12 Standard Test Method for High-Strain Dynamic Testing of Piles

D5780-95(2002) Standard Test Method for Individual Piles in Permafrost under Static Axial Compressive Load

D5882-07 Standard Test Method for Low Strain Integrity Testing of Piles

D6760 - 08Standard Test Method for Integrity Testing of Concrete Deep Foundations by Ultrasonic Crosshole Testing

D7383 - 10Standard Test Methods for Axial Compressive Force Pulse (Rapid) Testing of Deep Foundations

D7401 - 08Standard Test Methods for Laboratory Determination of Rock Anchor Capacities by Pull and Drop Tests

British / Euro Standards

BS 6349-2:2010. Maritime works Code of practice for the design of quay walls, jetties and dolphins

BS 8002: 1994 Earth retaining structures

BS 8004: 1986 British Standard Code of practice for Foundations

BS: 8081-1989 (Reprinted, Incorporated Amendment 1). Code of Practice for Ground Anchorages

BS EN 1536:2010. Execution of special geotechnical works. Bored piles

BS EN 1537: 2013 Execution of special geotechnical works. Ground anchors

BS EN 1538:2010 Execution of special geotechnical works. Diaphragm walls

BS EN 10248-1 1996 Hot rolled sheet piling Material standard

EN 10249:1996 Cold formed steel sheet piling

BS ISO 11886:2002. Building construction machinery and equipment. Pile driving and extracting equipment. Terminology and commercial specifications

BS EN 12063:1999. Execution of special geotechnical work. Sheet pile walls

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BS EN 12699:2001. Execution of special geotechnical work. Displacement piles

BS EN 12715:2000 Execution of special geotechnical work. Grouting

BS EN 12716: 2001 Execution of special geotechnical work Jet Grouting

BS EN 12794:2005. Precast concrete products. Foundation piles

BS EN 14199:2005. Execution of special geotechnical works. Micropiles

BIP 2205:2011. Concise Euro codes: Geotechnical design

BS EN 1997-1: 2004 Euro codes 7, Part 1 Geotechnical design. General rules

BS EN 1993-5:2007. Euro code 3. Design of steel structures Piling

BS EN 1997-1:2004. Euro code 7. Geotechnical design General rules

BS EN 1997-2:2004. Geotechnical design. General rules

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67

68

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