stonecutters bridge

8/11/2019 Stonecutters Bridge

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DESIGN AND CONSTRUCTION

OF STONECUTTERS BRIDGE,

HONG KONG

Naeem Hussain1and Steve Kite2

1Director, Ove Arup & Partners Hong Kong Ltd, Hong Kong2

Associate, Ove Arup & Partners Hong Kong Ltd, Hong Kong

ABSTRACT

Highways Department (HyD) of Hong Kong SAR is turning their idea for a world class bridge across

Rambler Channel at the entrance to Kwai Chung Container Port into reality. Due to the spectacular

location, HyD selected the concept for the bridge through an international design competition. The

competition took place in the first half of 2000 and the winning concept was a cable-stayed structure

with freestanding towers located between twin box girder decks. The 1018m main span is in steel,

while the four back spans each side are in concrete. The two towers stand on shore, providingunobstructed access to the busy container port with a minimum navigation headroom of 73.5m.

A number of modifications were introduced to the scheme during subsequent technical review.

Detailed design started in March 2002. Tender was called in August 2003 and returned in December

2003. The construction contract was awarded to the Maeda-Hitachi-Yokogawa-Hsin Chong Joint

Venture, and construction began in April 2004. At the end of 2007, the concrete back spans were

complete and the towers were well on their way to their final heights of 298m forming significant

landmarks against the backdrop of Hong Kong harbour. Major sections of the steel deck, fabricated in

China, are now on site, with the areas around each tower erected by a 4000T heavy lift strand jacking

scheme. The first of the prefabricated parallel wire stay cables have been erected. Completion of the

bridge is scheduled for 2009.

KEYWORDS

cable-stayed bridge, design competition, seismic design, typhoon wind loading, ship impact,

construction methods, steel fabrication, heavy lift scheme


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INTRODUCTION

An international design competition was organized by the Highways Department if the Government of

Hong Kong SAR in 2000 to establish a Reference Scheme (RS) for Stonecutters Bridge. It was won by

a joint venture of Halcrow and Flint & Neill from UK, Shanghai Municipal Design Institute from

China and Dissing & Weitling from Denmark. The bridge is the centrepiece of the new Route 8

expressway where the dual three lane road crosses the Rambler Channel at the entrance to the busy

container terminals (Figures 1 and 2). Route 8 will provide an alternative connection to the

international airport and better access to the new container terminal on Tsing Yi Island. The detailed

design assignment was won in a tender, based on technical and fee competition, by Arup in 2001 with

Cowi and BMT as principal sub-consultants.

BRIDGE LAYOUT

To provide a vertical clearance of 73.50m over the busy navigation channel, a cable-stay bridge with a

main span of 1018m, set to become the worlds second longest cable-stayed span, was chosen. The

total length is 1596m. The 53.30m wide main span deck comprises of two separate orthotropic steel

box girders connected by transverse cross girders spaced at 18m centres, to coincide with the stay-

cable anchorage spacing. The steel portion extends 49.75m beyond each tower so that the total length

Figure 1 : Location of Stonecutters Bridge

Figure 2 : Stonecutters Bridge A photomontage viewing from Lai Chi Kok


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Figure 5 : Steel deck - main span.

of steel deck is 1117.5m. The back span twin decks are in prestressed concrete and are monolithic

with the back span piers. Stay cables are in two planes arranged in a modified fan layout and attached

to the outside edges of the deck girders.

The mono-column towers are located in the opening between the twin decks, and are constructed in

concrete up to level +175m, and in composite construction with an outer stainless steel skin to +293m

and topped with a light feature to +298m. The base of the tower section is 24m by 18m, with circular

ends and straight sides. This tapers to become circular at deck level and upwards, with a diameter of

14m at deck level and 7m at the top. The back span piers are spaced at 79.75m, 70m, 70m and 69.25m.

The three intermediate piers have single rectangular tapered column shafts, whilst the end piers at the

interfaces with the adjoining viaducts are twin column portal structures. Both sides of the bridge are

on reclaimed land and foundations are large diameter bored piles taken down to rock. Pile lengths are

between 50m and 110m and diameters are between 2.2m and 2.8m (Figures 3, 4, 5, and 6).

Figure 3 : Elevation and plan of the bridge

Figure 4 : Concrete deck east back spans


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Design Memorandum

A project specific Design Memorandum was written to set out the rules and criteria for detaileddesign. The Design Memorandum referred to various codes of practice and design rules such as the

Hong Kong Structures Design Manual for Highways and Railways and BS5400, but it also defined

new loadings and rules for particular extreme events where the codes did not adequately deal with the

issues.

Limit State Design

In accordance with normal practice, design was carried out at serviceability limit state (SLS) and

ultimate limit state (ULS). At ULS a 5% chance of a loading event being exceeded within the 120-

year design life (exposure period) is permitted. This corresponds to a 2400-year return period. At

SLS a return period of 120 years is used resulting in a likelihood of exceedance of 63%.

Additionally, a third limit state for structural integrity (SILS) was defined for extreme events. The

design criteria chosen allow deformation and damage to occur, but no destruction of key elements, and

the bridge should remain operational for emergency traffic. After such an event, significant repair

might be required and the bridge could be closed to allow this to take place. A 2% likelihood of

exceeding the condition with a corresponding 6000-year return period is deemed appropriate.

DESIGN FOR SEISMIC ACTION

Short span bridges in Hong Kong have traditionally been designed for a 0.05g nominal lateral load.

For long period structures this is too onerous and a more sophisticated approach using responsespectra is required.

Ground Motion Hazard

Ground motion can be quantified in several ways. From historical records observations of damage can

determine a measure of intensity, although descriptive records are obviously subjective. Where the

data is available, peak acceleration, velocity or displacement values give a more exact measure. The

frequency content of motions represented in response spectra gives a more complete picture.

The key information required for a seismic hazard assessment includes the proximity, and the activity

rate of potential seismic sources, and the attenuation between the source and the point of interest.

Clearly the further away from the source the less intense the ground motions are at the site.

Figure 6 : Foundations East side


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Probabilistic Seismic Hazard Assessment (PSHA)

A PSHA was carried out to determine ground motions having the desired annual probabilities of being

exceeded. The observed seismicity in the South China region came from historical data (measure of

intensity) and more recent recorded data compiled by several agencies. The onshore data is much

more complete than the offshore records, but allowing for the different completeness a coherent and

consistent source model can be determined (Free, Pappin & Koo, 2004). The source model for the

500km range from Hong Kong shown in Figure 7 was developed. Within each area a uniform activity

rate is assumed. A constant relationship was used to model the relative activity rates for different

magnitudes. Taiwan was treated as a separate source model and it had a significant contribution to the

SLS long period motion.

Site Response

The effect of the local soil profile was taken into account by using a one-dimensional non-linear

dynamic site response analysis to determine the design spectra at ground level. Site response effects

can amplify the underlying ground motions, in some cases considerably. In-situ shear wave velocity

tests were used to determine the soil input parameters for these analyses. The spectral results from thetime history analyses were enveloped into an easily defined spectrum for each limit state return period.

See Table 1, and Figures 8 and 9.

TABLE 1SPECTRALACCELERATIONSFORDIFFERENTLIMITSTATES

Spectral Acceleration (m/s2)

for each Period (s)Limit

State

Return

Period

(years)

Probability of

being exceeded

in 120 years

Peak

Acceleration

(m/s) 0.1 0.5 1.0 2.0 5.0 10

SLS 120 63 % 0.70 1.75 1.75 0.88 0.44 0.07 0.02

ULS 2400 5 % 2.40 6.00 6.00 3.00 1.50 0.24 0.06SILS 6000 2 % 3.15 7.88 7.88 3.94 1.97 0.32 0.08

Figure 7 : Earthquakes within 500km of Hong Kong

1

2

3

4

5

6

7

1

2

3

4

5

6

7


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Liquefaction Scenarios

Liquefaction occurs when the strength and stiffness of a saturated soil is reduced because of the

increase in pore water pressure under rapid loading, such as a seismic event. The probability of

liquefaction was studied and considered to be likely at SILS. Lower values of soil shear modulus over

the affected depth where liquefaction may occur were therefore used in the pile group analysis for the

SILS condition.

Non-synchronous Ground Motions

When considering a structure of significant size such as Stonecutters Bridge, the effects at one tower

may be different from those at the other tower, 1.0 kilometre away. The relative movement between

supports is assessed by examining a displacement time history analysis to determine the maximum

differential horizontal displacements. Simple static rules can then be used to add this effect into the

results from the standard spectral analysis.

DESIGN FOR WIND ACTION

Hong Kong is affected by typhoon winds and the bridge must therefore be able to withstand the

associated extreme wind loading. In order to establish the wind climate and wind loading model

suitable for the dynamic wind load assessment of the bridge, wind tunnel tests and analyses of newand existing data were undertaken.

Buffeting calculations allow assessment of the actions on a flexible structure arising from the

interaction between gusty winds and the dynamics of the structure itself. In order to carry out such

calculations, not only the wind speed profile must be known but also the turbulence properties in the

space occupied by the bridge structure: i) the turbulence intensity profile which is a function of the

height above ground level, ii) the spectral distribution of turbulent velocity fluctuations which is a

function of frequency, and iii) the coherence which is a function of frequency and spatial distance

between points.

A large amount of historical wind data was available from the Hong Kong Observatory at Waglan

Island, south east of Hong Kong Island. However, the bridge is located within Hong Kong Harbour

and partly surrounded by hills and high-rise urban development affecting the wind characteristics.

Figure 8 : Spectral accelerations Figure 9 : Calculated site spectra

2400 year Velocity Response Spectrum 5% damping


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Further existing data was also made available from Tsing Ma and Kap Shui Mun Bridges, which have

locations more akin to Stonecutters Bridge. Although the analysis of this data provided valuable

estimates of the mountainous/urban exposure turbulence to be expected, it could not be considered a

priori to be accurate at the new bridge site. Therefore terrain model testing at 1:1500 scale of the

bridge site, including Tsing Yi Island and the Tsing Ma/Kap Shui Mun area, was carried out at

Monash University in Australia (Figure 10), as well as wind turbulence measurements on site at the

location of the eastern end of Stonecutters Bridge. Results from the wind tunnel were then correlated

with site measurements.

Ocean exposure is characterised by high wind speeds and low turbulence. The characteristic 1-hour

design mean wind speed at deck level equals 52m/sec for ocean exposure from south westerly

directions. When the wind is approaching from the mountainous/urban terrain the maximum wind

speed is lower but the level of turbulence is high. The 1-hour design mean wind speed at deck level

equals 42m/sec for this exposure from westerly to south easterly directions. Refer to the location plan

in Figure 1. The high turbulence wind has generally governed the design.

The design turbulence intensity functions at height z above ground have been adopted as shown below.

19.0

10

10175.0

)(

==

zzVI uu

29.0

10

10437.0

)(

==

zzVI uu

DESIGN DEVELOPMENT

The development of the RS was carried out to improve the constructability, the structural/aerodynamicperformance and/or maintainability/durability of the bridge, but preserving the appearance established

through the design competition. Some keys improvements are described below.

Figure 10 : Terrain model in wind tunnel

For approach wind from open ocean fetch from south-

westerly direction.

For approach wind over complex hilly or built-up terrain

from westerly to south easterly direction.


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Articulation

The RS included a monolithic joint between deck and tower. This was replaced by transverse bearings

for lateral loadings and longitudinal hydraulic buffers for short term wind and seismic loads as shown

in Figure 11. The benefits of the change are:

Reduced restraining forces on the tower from temperature variations for the bridge in-service.

Reduced torsional moment in the tower during construction (cantilevering of the main span) as

well as in-service, which was very large under buffeting wind effects.

Elimination of a hard-point in the deck for bending about a horizontal axis to avoid a

combination of high bi-axial bending and compression.

Tower

Circular shaped towers are susceptible to vortex shedding induced vibrations, which can occur at low

wind speeds and induce vibrations of the stay cables due to linear resonance and parametric excitation.

The upper part of the RS tower was in steel and the lateral tower frequencies matched the frequencies

of some of the stay cables. An aero-elastic wind tunnel test at a 1:100 scale at the Velux laboratory in

Denmark showed large sinusoidal motion of the tower with a peak response of up to 0.45m. A time-

history analysis, run over 400 seconds, showed that this could induce a peak amplitude of stay cable

motion of up to 8.35m due to linear resonance vibration. A composite tower with 800mm thick

concrete and 20mm steel skin (Figure 12) overcame the problem by making the frequency of vibration

of the tower outside the range of frequency of the stay cable vibrations, and also reducing the tower

peak amplitude deflection to 0.07m.

Figure 11 : Transverse bearings and hydraulic buffers at towers

CABLE ANCHOR BOX 20mm STAINLESS

STEEL SKIN

800mm CONCRETE

WALL

Figure 12 : Details of the upper tower


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The cross girder tendons are all placed at maximum eccentricity at the bottom of the section, and in

order to stress them fully prior to the deck being supported by the stay cables, temporary prestress

above deck level is required. This temporary prestress was tuned to lock in additional bending to the

cross girders. When the stay cables are later attached and tensioned at the deck edges, the temporary

transverse prestress will be released and a beneficial torsion is induced in the two longitudinal girders

(Figure 14).

The longitudinal prestress is carried out when all deck units are completed and connected to the pier

crossheads. The longitudinal prestress consists mainly of external tendons inside the boxes with

tendon deviation at the cross girder walls. The prestress demand for flexure reduces with the

increasing compression from the inclined stay cables towards the towers. The interface between

concrete and steel is prestressed such that the joint is in compression over the entire area for SLS loads

and that sufficient compression is present to transfer the shear forces safely at ULS.

Design of Steel Decks

The steel boxes have orthotropic deck plates, the design of which is governed by fatigue loading on

the bridge. The bridge is located in a sub-tropical climate with afternoon summer time temperaturesfrequently above 30C. The reduction in stiffness of asphalt surfacing at high temperatures means that

the benefit of the surfacing acting compositely with the deck plate to reduce local stresses will be

limited. Furthermore, the traffic crossing the bridge is expected to contain an unusually high number

of heavy goods vehicles due to the bridge linking to the Kwai Chung container terminal. It is

estimated that 42% of the total traffic will be heavy goods vehicles. To cope with this intense fatigue

loading, without beneficial composite action with the surfacing, the orthotropic steel decks have been

designed with an 18mm thick deck plate together with 325mm deep, 9mm thick, trough stiffeners.

The distance between the diaphragms is 3.8m in general, and 2.8m at the cross girders. This results in

a relatively heavy and stiff top flange, which will also be advantageous in extending the life of the

60mm thick mastic asphalt surfacing.

Design of Stay Cables

The stay cables are the prefabricated parallel wire type with 7mm diameter galvanised wires and

extruded outer HDPE sheathing. The tensile strength of the wires is 1770MPa and the allowable stress

at service load is 769MPa. The stay cables are very compact, with outer diameters varying from

113mm (163 wires) near the towers to 192mm (499 wires) towards the end of the back spans. The

longest cable is 540m long and weighs about 70 tonnes.

Full scale wind tunnel tests were carried out on stay cables of various diameters to confirm the drag

coefficient used in the design and to investigate the effect of various surface profiles to counteract

rain-wind induced vibrations. Hydraulic internal dampers will be installed at deck stay anchorage

tubes and rubber dampers at the tower anchorage tubes.

Figure 14 : Schematic prestress layout at cross girders

TEMPORARY PRESTRESS


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Design of Foundations

Accidental Loads due to Ship Impact

The tower foundations are located approximately 10m from seawalls on both sides of the Rambler

Channel. Given the close proximity, account has been taken in the design for impact loading induced

by a ship collision with the seawall. A series of centrifuge tests were carried out (Lee & Peiris, 2004)

to model the effect of a 155,000 tonnes container ship impacting the seawall at a speed of 6 knots.

From the results of the test and correlation of the pressure measurements with elastic numerical 3D

models, the force exerted by the impact at the front face of the bridge foundation has been determined

(Figures 15 and 16).

Foundation Design

The foundations have been designed using an iterative process to achieve compatibility between the

superstructure and the substructure of the bridge. The pile foundations are designed to accommodate

additional negative skin friction loads resulting from down-drag of the soil caused by the ongoing

long-term settlement of the reclaimed ground. Allowable bearing pressures vary from 3.0MPa for

moderately decomposed rock to 7.5MPa for fresh to slightly decomposed strong rock. In order to

achieve optimum design of the bored piles, the bases of the piles are enlarged to form bell-outs to

increase their bearing capacity such that they are of similar magnitudes as the structural compression

capacities of the pile shafts.

CONSTRUCTION

The construction contract was awarded to a Joint Venture (MHYHJV) led by a Japanese contractor,

Maeda, together with two other Japanese contractors, Hitachi and Yokogawa, and a local Hong Kong

contractor, Hsin Chong.

The handing over of the construction site was staggered, with the Contractor getting immediate

possession of the eastern side in April 2004 while the western side was handed over in October 2004.

Foundations

Piling

The piles were constructed to tight positional and verticality tolerances using full depth temporary

steel casings installed with an oscillator. For the longest piles some of the casings were sacrificial. A

Figure 15 : Ship impact model in centrifuge Figure 16 : 3D computer model of impact


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grab was used to excavate the sand, followed by rotary core drilling to form bell-outs in rock. The

bellouts were not cased, so there was potential for instability and void loss. This problem was

overcome by grouting the zones where necessary, and then coring though the grout, leaving an

annulus of stable material.

Pile caps

Constructing the pile caps in the permeable sand next to the sea required careful design of the sheet

pile cofferdams and dewatering systems. The back span caps, typically 19m by 11m by 4m thick,

were cast in a single pour. To control differential temperatures, insulation was provided to retard heat

dissipation (Figure 17). Each tower cofferdam was a 38m by 50m by 10m deep excavation and had

three layers of steel struts which were incorporated into the caps (Figure 18). Concrete pours typically

1m thick were used to form the 8m thick caps, with additional reinforcement provided at each layer to

control thermal cracking.

Concrete Back Spans

Pier Shafts

The intermediate pier shafts are between 60 and 65m tall, with hollow box sections tapering from

12.5m to 10m wide, having a constant thickness of 4m. Walls are either 600mm or 1m thick. They

were constructed with 60MPa concrete using a hydraulic climbing form system from VSL. Four

pockets were cast into the outer face at each pour to provide the support points for the climbform.

With typical pour heights of 4m, a cycle time for construction of each lift of 6 days was achieved, withconcrete finishing works undertaken from trailing platforms which hung below the main working

platforms.

The end portal pier shafts have a similar form and were constructed using similar techniques.

Pier Cross Heads

At each intermediate pier, the monolithic cross head is formed by in-situ cantilever construction. The

integration of the deep section (9m at the root) with the curved soffits of the longitudinal decks made

for a complicated geometry, as shown in Figure 19, so the pours are split into manageable sizes. Steel

temporary works trusses cantilevering from the pier shaft provided the support in the temporary

construction stage before the concrete has gained the required strength.

Figure 17 : Back span pile cap Figure 18 : Tower pile cap


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Concrete Deck

Falsework System

The contractor elected the use of precast columns on bored pile foundations to provide the required

temporary support system which carries the weight of the in-situ concrete decks until the stay cables

are installed. Two columns per cross girder up to 60m tall are spaced at 35.8m to suit the web

locations. This system provides a relatively stiff support, which minimises the deflections that occur

when the deck segments are cast, and also minimises uncertainty of those deflections, hence reducing

the adjustments required to correct the levels. The 50MPa concrete columns are hollow sections, 2.5m

square, with typical wall thicknesses of 250mm. Vertical prestressing bars are coupled together and

stressed down onto the segments at each of three 20m high modules. Steel truss members brace the

columns together in both the longitudinal and transverse directions, and plan bracing is also introduced

at the top of each module to form structures rigid enough to carry the heavy decks even in the event of

a full typhoon.

Level adjustment is provided at the top of each temporary column, with four 500T jacks, which could

be used to move elements of the decks up or down as necessary. This ensures the correct alignment of

the independent cross girders, prior to connecting them to each other and then to the pier cross heads,

even if the falsework deflections are different from the predicted values.

Cross Girders

The 4m wide concrete cross girders were cast on stiff steel trusses, spanning the 35.8m between the

falsework columns, and wing trusses for the areas outboard of the columns. The area on top of the

trusses was decked out, with scaffolding on top of this to form the curved shape of the soffit.

The contractor has modified the envisaged arrangement of temporary transverse prestress slightly to

give better clearance for access along the top of the deck. The pairs of horizontal cables, providing a

force of 5MN, are located 4.5m above the top slab. Additional transverse prestress tendons are

provided in the top slab in the region above the falsework columns to counteract the hogging moment.

Longitudinal Girders

Steel falsework trusses were hung from the constructed cross girders and cross heads to support

casting the longitudinal box girders in each bay. Similar decking as used for the cross girders was

placed on these trusses to form working platforms, and then scaffolding provided to form curved soffit

shape. Due to larger pours and complex shapes, a three staged pour is used the bottom slab first,

Figure 19 : Cutaway section of typical pier cross head


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then the webs, lastly the top slab with careful evaluation of the locked-in stresses this induces to

ensure that they can be accommodated by the design.

Lower Towers

The complex shape of the tapering lower towers from ground level to +175m was formed using a

climbing formwork system provided by subcontractor Cantilever. 10 individual panels carried the

plywood shutters (Figure 21). Strips were cut off the edges each time the form was lifted to reduce the

perimeter length for each pour. The high quality plywood had to be durable enough for the repeated

pours, but also flexible enough to be bent into an arc with an ever decreasing radius.

The climbing operation to raise the form in

preparation for the next pour is controlled by 10

pairs of screw jacks, supported on the top of the

previous construction joint. Steering the form to

maintain the correct alignment was a delicateoperation with coordination of all jacks required

to keep the form level and eliminate any twisting.

A cycle time of 7 days was achieved for the

typically 4m high pours, with concrete finishing

works undertaken from trailing platforms which

hung below the main working platforms.

Figure 21 : Lower Tower construction

Figure 20 : Back Span construction in progress


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Upper Towers

In each tower, 32 stainless steel sections make up the tapering outer skin. The lowest 3 sets of stay

cables anchor in corbels on the inside face of the concrete wall, whereas the remaining stay cables

anchor within a steel box made of 25 carbon steel sections forming the core of the tower between

levels +195m and +280m.

Steelwork Fabrication

Duplex stainless steel, grade 1.4462, is adopted for the steel sections, as it provides a very high level

of corrosion resistance in the marine atmosphere and avoids the need for painting of carbon steel in

areas which are extremely difficult to access. Steel plate was supplied by Outokumpu from Sweden to

the fabrication yard in Guangdong province, China. Each segment is fabricated in 2 halves from

20mm thick plate, with 25mm thick stiffening flanges top and bottom, and intermediate stiffening

rings. Vertical bolted splice connections join the halves together, and horizontal bolted splices

connect segments together at the flanges. 300mm long shear studs on the back face of the skin form

the composite connection into the concrete wall.

Anchor boxes, as shown in Figure 22, have plate thicknesses up to 40mm thick, which require

considerable heat control to minimise distortions during welding. Careful planning of the assembly

sequence was also required to limit the areas where access for welding is difficult. Shear studs on the

end faces and corners of the anchor boxes provide composite action with the concrete wall.

After the match fabrication of adjacent segments, the skins are fitted around the central anchor boxes

in a rolling trial assembly operation of three segments to ensure the correct geometry and verticality.

(Figure 23). This process is vital to ensure accurate fit between segments, and to maintain their

vertical alignment. During erection on site, the geometry achieved during the trial is repeated, with

little or no room for adjustment, so making certain that it is within tolerance at the fabrication stage isessential. The relative geometry between the skins and the anchor boxes dictates the correct alignment

and position of the stay cable anchorage points.

Figure 22 : Fabricated anchor box

Figure 23 : Trial assembly of 3 segments

of tower steelwork


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The final operation in fabrication is the shot-peening process to achieve the desired surface finish on

the outside face of the stainless steel skins. A mixture of fine glass bead and aluminium oxide was

blasted into the surface to produce a slightly textured finish, rough enough not to reflect too much

sunlight which could dazzle drivers, but smooth enough to avoid the accumulation of excessive

amounts of dirt.

Erection

Tower cranes are used to erect the stack of steelwork on top of the lower towers, reproducing the same

geometry as achieved in the trial assembly. The verticality of the first skin was crucial in achieving

the correct alignment of the remainder of each upper tower. Therefore, at each tower, the first 2 skins

were positioned and extensive surveys carried out prior to finalising the levelling and grouting the

interface between the lower tower and upper tower.

The heaviest half skin, including temporary stiffening and a lifting beam, weighed 24T which dictated

the type of tower crane used. The lightest half is only 9T, and this is the one to be lifted to the highest

elevation, where stronger winds are more likely. Concerns over controlling the behaviour in strongwinds led to strict limits on the allowable wind speed for lifting to take place. Real-time wind

measurements from the top of the tower cranes are transmitted to the site team. Anchor boxes weigh

between 18T to 15T with shapes much less likely to suffer adverse effects in the wind.

In general, the erection of skin sections advances a full cycle ahead of the concreting. Reinforcement

fixing is a relatively difficult operation due the small working area available inside the skin and the

careful integration of steel reinforcement with shear studs of both the skin and the anchor boxes. The

cycle sequence ensures that whenever concrete is being placed the shape of the skin is held in position

by the skin above.

At the end of each cycle a survey of each stay cable anchorage point is undertaken so that anytolerances can be taken into account in determining the stay cable installation lengths.

Figure 25 : Segment assembly of deck panelsFigure 24 : Upper Tower construction


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Steel Deck

Fabrication and Assembly

Steel deck panels are fabricated at a facility in North Eastern China. The plates are cut to shape and

stiffeners are welded on to create sub-assemblies such as deck plates, bottom panels, diaphragms and

cross girder panels. An initial mock-up segment was constructed to highlight where difficulties would

be expected in fabrication and to confirm the fit-up geometry. The lessons learnt have been applied to

both panel fabrication and subsequent assembly.

The panels are then shipped to an assembly yard in Guangdong province, Southern China, where

assembly takes place on two production lines (Figure 25). Match fabrication to ensure a consistent

cross section shape and the correct segment alignment is crucial to ensure that welding segments

together on site proceeds without problems. Blasting and painting is the final operation before

shipment to Hong Kong by barge.

Heavy Lifting Scheme

The 88m length of steel deck above land was originally envisaged to be constructed on high level

falsework, similar to that used for the concrete back spans. However, the contractor elected to use a

heavy lifting scheme in which the two longitudinal girders are each assembled at ground level and

then simultaneously strand jacked 70m into their final positions. The total lifting load is 4000T,

including the weights of lifting frames to be used later for main span segment erection.

Due to the shape of the tower, which is wider at the base than at deck level, the two decks are 12m

further apart at ground level, and are slid transversely once at high level. A 2m longitudinal slide is

also necessary to place the decks onto a temporary interface truss before lifting and welding the

connecting cross girders, and casting the 2m section to stitch the steel and concrete decks together.

Figure 26 : 4000T Heavy Lift at the East Tower in progress


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Each deck was suspended on four lifting points, with 88% of the load carried at temporary brackets

attached to the tower. The remaining load was carried via temporary lifting beams cantilevering from

the concrete back span decks. Using the permanent works to support the lifting equipment eliminated

additional foundations and lifting pylons, whilst allowing the assembly work to be carried out at

ground level. However, a high degree of integration between temporary and permanent works is

required, which demands careful planning in order to minimise disruption to the on-going construction

programme.

Main Span Erection

53 main span segments will later be erected by lifting frames in progressive cantilever erection out

over the Rambler Channel. 18m long segments weighing up to 600T will be lifted with high capacity

winches from a dynamic positioning barge. A 200m square work zone, and an 8 hour window for

lifting operations will ensure minimum disruption to shipping. Jacking to match the geometry of the

new segment, supported on winch ropes, to the existing geometry of the cantilever, supported at its

outer edges by the stay cables, will be required to allow the segment joint site welding to take place.

Stay Cables

Nippon Steel are supplying prefabricated parallel wire cables, which arrive on site from China on large

diameter drums. The first stays are unreeled at ground level and lifted up into place to enable the

sockets to be installed into the tower and the deck (Figures 27, 28). Once main span cantilevering

commences, unreeling will take place along the deck. Stressing is performed underneath the deck,

from temporary hanging access platforms to support the heavy jacking equipment. Temporary

measures to control cable vibrations are used until installation of internal hydraulic dampers to achieve

the required level of damping in the permanent condition.

Figure 27 : Stay cable unreeling and lifting Figure 28 : First stay cable installed


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REFERENCES

Free, Pappin & Koo (2004). Hazard Assessment in a Moderate Seismicity Region, Hong Kong, 13th

World Conference on Earthquake Engineering, Vancouver, B.C., Canada, August 1-6, 2004, Paper No.

1659.

Lee & Peiris (2004). Modelling of Ship Impact on a Bridge Foundation,IABSE Symposium

Shanghai 2004: Metropolitan Habitats and Infrastructure, IABSE Report, Volume 88, 2004.

stonecutters bridge

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