bridge design manual-texas department of transportation

Bridge Design Manual

December 2001© by Texas Department of Transportation

(512) 416-2055 all rights reserved

Manual Notice 2001-1

To: Districts, Divisions and Offices

From: Steven E. Simmons, P.E.Deputy Executive Director

Manual: Bridge Design Manual

Effective Date: December 1, 2001

Purpose

This manual provides policies and guidelines set forth by TxDOT regarding the design ofbridges. It has been developed to help bridge designers working on TxDOT projects.

Instructions

This is a new manual containing new and significantly reorganized material. It supersedesthe 1990 Bridge Design Guide and the 1990 Bridge Design Examples, both first editions.

Contents

The manual contains ten chapters – Organizational Overview, TxDOT and Bridge Design,Design Specifications, Geometric Restraints, Preliminary Considerations, General DesignControls, Superstructure Design, Substructure Design, Special Designs, and FoundationDesign. The manual also has four appendices.

Contact

For more information regarding any chapter or section in this manual, please contact theDesign Section of the Bridge Division.

Bridge Design Manual 1-1 TxDOT 12/2001

Chapter 1 Organizational Overview

Contents:Section 1 — This Manual .....................................................................................................1-2

Section 2 — Evolution of the TxDOT Bridge Division .......................................................1-4

Chapter 1 — Organizational Overview Section 1 — This Manual


Section 1 This Manual

Overview

This manual was developed to provide bridge designers working on Texas Department ofTransportation (TxDOT) projects with the policies and guidelines set forth by TxDOTregarding the design of bridges.

Its purpose is to improve the bridge design and detailing process by promoting uniformityamong bridge designers working on TxDOT projects.

This manual is subject to revision as conditions, experience, or research data warrant.Changes will be issued by Manual Notice. Changes may be in the form of new sheets to beadded, revised sheets to replace superceded ones, or sheets to be deleted.

The manual is not intended to be a complete substitute for engineering experience,knowledge, or judgment. Special situations may arise that appear to call for variation fromthe policy requirements herein. Such variation will be subject to approval of theadministration of the TxDOT Bridge Division.

Direct any questions or comments on the content of the manual to the Director of the DesignSection of the Bridge Division, Texas Department of Transportation.

Bridge Design Manual Format

The manual begins with this overview of the manual and a description of the evolution ofthe TxDOT Bridge Division. The chapters that follow include information on TxDOTDivisions/Sections, design specifications, geometric restraints, preliminary considerations,general design controls, superstructure design, substructure design, special designs, andfoundation design. The following paragraphs briefly discuss these chapters:

Chapter 2 presents a description of the TxDOT Divisions/Sections primarily involved inbridge design, planning, construction, and maintenance and provides descriptions of theresponsibilities of the TxDOT Bridge Division’s Bridge Design Section.

Chapter 3 lists and briefly describes the governing design specifications, or “Rule Books,”involved in bridge design. The chapter includes information on mandatory specifications,guide specifications, and industry recommendations.

Chapter 4 discusses the common roadway geometric restraints inherent in bridge design.Bridge widths, span lengths, clearances, and alignment are discussed. A section on theconstraints involved during stage construction is also included.

Chapter 5 presents some common aspects a designer/planner must consider during thepreliminary planning and design process. These aspects include materials, structure type,

Chapter 1 — Organizational Overview Section 1 — This Manual


economics, and aesthetics. A discussion on bridge railing and use of corrosion protection isalso included.

Chapter 6 discusses in greater detail some of the more common design specificationsinvolved during the design of a bridge, giving the designer additional information on theapplication and usage of common design specifications and criteria.

Chapter 7 presents design criteria and design guidance for the most commonly usedsuperstructure types, including cast-in-place, precast, and steel superstructures. Backgroundinformation on the development of each superstructure type is also included.

Chapter 8 presents design criteria and design guidance for the most commonly usedsubstructure items, including caps, columns, and foundations. Background information onthe development of some of these items is also included.

Chapter 9 presents design criteria and design guidance for designs that inherently involveunique aspects, culverts and drainage, bridge appurtenances, sign bridges, and somecommon bridge items. Background information on the development of some of thesedesigns is also included.

Chapter 10 discusses in greater detail the relationship between structural design andgeotechnical design. Some guidance on bridge foundation designs and retaining walldesigns is included, as well as background information on the development of some of theseitems.

Chapter 1 — Organizational OverviewSection 2 — Evolution of the TxDOT Bridge

Division


Section 2 Evolution of the TxDOT Bridge Division

Origin of the TxDOT Bridge Division

The Texas Highway Department was established in 1917 and is responsible to the Governorof Texas to design, construct, and maintain an adequate system of highways in the state. In1918, a Bridge Office was created with the primary responsibility of preparing standarddesigns and drawings in an attempt to bring some uniformity to the bridges beingconstructed by the counties. The Bridge Division appeared in 1928, retaining bridge designas a big part of its mission.

The Bridge Division continued to maintain standards and design non-standard bridges. Intime, advance planning, railroad negotiations, and plan review capabilities were developed.Construction management was provided for some of the more complicated structures.

1940s and 1950s

Activities were curtailed during the war years, but in the late 1940s and 1950s increaseddemand for improved infrastructure produced a large volume of expressways, for whichspecial design offices were established in the affected cities. Some of the groups adoptedtheir own design and detailing standards.

When welding began to replace rivets for field splices in steel beams and girders in the early1950s, the Bridge Division sent qualified welders to the larger projects to help with qualityassurance and quality control.

In the middle 1950s the Bridge Division, with the cooperation of precast manufacturers,developed a group of standard pretensioned concrete beams, which quickly proved to be themost economical way to construct medium-span length bridges.

When the Interstate Highway System was inaugurated in the middle 1950s, the designworkload increased dramatically and has remained generally good to date. Between theexpressway offices, district design groups, and the Bridge Division, plan preparation washandled for several years with a minimum of help from consulting engineers.

Recent Years

In the early 1980s, consulting engineers began to do a significant portion of the highwayplans and a somewhat smaller portion of the bridge plans. After a period of reducedactivity, consulting engineers are now preparing a significant portion of highway and bridgeplans.

Meanwhile, the use of TxDOT “bridge standards” has become more uniform, as manydistrict design groups have abandoned their own plan preparation activities. Currently, the

Chapter 1 — Organizational OverviewSection 2 — Evolution of the TxDOT Bridge

Division


Bridge Division continues to prepare its share of structure plans while attending to agrowing number of non-engineering responsibilities.


Chapter 2 TxDOT and Bridge Design

Contents:Section 1 — Coordinating with TxDOT Divisions and Sections .........................................2-2

Section 2 — Primary Responsibilities of the Bridge Design Section...................................2-3

Section 3 — Coordination Responsibilities of the Bridge Design Section...........................2-7

Section 4 — Contractive Responsibilities of the Bridge Design Section ...........................2-10

Chapter 2 — TxDOT and Bridge DesignSection 1 — Coordinating with TxDOT Divisions

and Sections


Section 1 Coordinating with TxDOT Divisions and Sections

Overview

Before a bridge is designed, critical preliminary functions described in theBridge Project Development Manual must be completed. The planning and design of abridge project involves several divisions and sections within the Texas Department ofTransportation (TxDOT). Some of the contributing entities are:

♦ Bridge Division

♦ Design Division, Field Coordination Section

♦ Transportation Planning and Programming Division

♦ Traffic Operations Division, Railroad Section

♦ Environmental Affairs Division

♦ Construction Division, Materials Section

♦ Maintenance Division, Maintenance Operations Section

Within the Bridge Division, the Bridge Design Section is responsible for functions thatinclude engineering and non-engineering aspects of bridge design. These responsibilitiesare discussed in this chapter.

Chapter 2 — TxDOT and Bridge DesignSection 2 — Primary Responsibilities of the Bridge

Design Section


Section 2 Primary Responsibilities of the Bridge Design Section

Overview

The primary responsibilities of the Bridge Design Section are structural design and thepreparation of working drawings or plans. These primary responsibilities involve aprocedure that begins with a concept to construct a highway facility and concludes with thesubmission of finalized plans, specifications, and estimates (PS&E). The procedure includesmany steps and it is important to know these steps to fully grasp the responsibilities of theBridge Design Section. Generally, the procedure for preparation of plans by the BridgeDesign Section is as follows.

Consultation. Consultation between the Bridge Design Section, bridge project developmentmanager, and the district design engineer, district bridge engineer, and/or area engineershould precede determination of structure type and scheduling of the letting.

Note: This is a good time to make a preliminary decision about who will prepare the bridgeplans.

Preliminary Bridge Layouts. The area engineer or the project’s designated consultingengineer prepare preliminary bridge layouts. These layouts are usually complete withgeometric controls, type, size, length of spans, hydraulic data, required clearances, soil testboring data, classification of highway, and projected traffic. At this time, type of foundationshould be proposed and conveyance of water through stream crossings and scour analysisshould be addressed and coordinated with the Hydraulics Section.

Note: Area engineers and consulting engineers are encouraged to contact the GeotechnicalBranch for advice if there is any question regarding the proper foundation.

The layouts are sent to the bridge project development manager who will forward them tothe Design Division, the Federal Highway Administration (FHWA) on federal oversightprojects, or other agencies that may exercise review authority.

Bridge Plan Preparation. When approval has been secured from all the appropriateagencies, timing for the plan work is re-negotiated with the district and the job of bridgeplan preparation is given to the bridge design engineer or to the consultant.

The following steps apply to the Bridge Design Section, but the routine for consultants willbe similar.

Note: If consultants are unsure about the current design or detailing standards for an item,they are encouraged to contact the director of the Bridge Design Section.

♦ The director of the Bridge Design Section assigns the work to a Design Groupaccording to its particular expertise in that type of design, and primarily on its ability tocomplete the plans in the required length of time.


Design Section


♦ The Design Group leader schedules engineering and detailing work for the jobaccording to the target completion date and the Design Group’s other commitments.

♦ The Geotechnical Branch is contacted early if there is any doubt about the foundationtype. When foundation loads have been determined, the Geotechnical Branch will beasked to establish founding elevations.

♦ When the plans are complete, the bridge project manager sends prints to the district.

♦ Reviewed prints are returned to the bridge project manager and the Design Groupmakes any revisions required by the district review.

♦ Originals, including all reproducible standard drawings are sent to the district.

♦ Project plans, specifications, and estimates are sent to the Design Division. Anyrevision required by the pre-letting review are made by the Design Group.

♦ After letting, optional designs for prestressed concrete beams are submitted by thefabricator and checked early. Other shop drawings follow.Note: Consulting engineers should be aware that shop drawings require a significantamount of checking time. They should budget their labor accordingly.

♦ Occasionally, bridge members that were fabricated beyond specification tolerances arereviewed by the Bridge Design Section for structural adequacy when properly repaired.Also, occasionally, construction problems arise that require review for structuraladequacy.

♦ When the project is finished, geometric calculations are discarded and design notes areassembled and filed.

Structural Design

Structural design involves selection of appropriate materials, systems, and details for thestructure and performing calculations of stress and strain in each component caused by theprescribed loading. The following items give examples of potential areas of consideration.

Stream Crossings. Stream crossings carry highway traffic over creeks, rivers, bayous,channels, and bays. Hydraulic considerations are usually involved. Clearances for marinetraffic may be required.

Grade Separation Structures. Grade separation structures occur where one roadway mustcross over another. Clearance for the overpassed traffic is critical. Highway over highwayseparations are called overpasses if the project highway passes over; otherwise they arecalled underpasses.

Railroad Underpasses. Railroad underpasses are where a highway passes under anintersecting railroad. Design is under strict control by the railroad companies.

Miscellaneous Structures. Miscellaneous structures include sign bridges, illuminationpoles, traffic signal supports, pedestrian overpasses, utility bridges, movable bridges, fendersystems, ferry boat landings, and radio towers.


Design Section


Culverts and End Treatment. Culverts carry storm water under highways. The larger ofthese culverts are reinforced concrete boxes. Most culverts are constructed from standarddrawings. Under some conditions Safety-End Treatment is required to protect errantvehicles. These conditions are outlined in the TxDOT Roadway Design Manual.

Retaining Walls. Some types of retaining walls require structural design. Mechanicallystabilized earth (MSE) wall designs require mostly geotechnical considerations. Detailedplans for MSE walls are prepared by the successful wall supplier.

Geotechnical Design. A special group within the Bridge Technical Services Sectionperforms geotechnical design.

Preparation and Approval of Working Drawings

Preparation and approval of working drawings involve assuring that the variousrequirements of the design are shown on plan-size (22 x 34 in.) or half-size (11 x 17 in.)drawings as completely and accurately as necessary to allow the structure to be builtaccording to the design. Half-size sheets are preferred by the department. The use of full-size sheets is being phased out.

Bridge Plans. Drawings for construction projects must contain accurate quantities of thevarious items of work so that the contractor can be adequately reimbursed according to theunit bid prices.

Bridge Standards. Bridge standards are maintained by the Bridge Standards Branch of theTechnical Services Section. Standard drawings contain often-used systems and details thatcan be used in bridge plans without modification. Standards are indexed on the mainTxDOT website under Business/TxDOT CAD Standard Plan Files. (Seehttp://www.dot.state.tx.us/insdtdot/orgchart/cmd/cserve/standard/disclaim.htm.) The websitecontains instructions about the use of the graphics files.

Preliminary Bridge Layouts. Preliminary bridge layouts are reviewed and approved by theBridge Design Section. Layouts are initiated by the district, sometimes with assistance by aconsulting engineer. See the Bridge Project Development Manual for the submittal process.

Note: For major structures, coordinate with the Bridge Design Section through the projectdevelopment manager as early as possible.

PS&E. Plans, specifications, and estimates contain structural details prepared by thedistrict, consulting engineers, or the Bridge Design Section. In any case a review isnecessary to determine that the structure is safe and reasonably economical.

Shop Drawings. The Field Operations Section of the Bridge Division coordinates andchecks shop drawings. The Bridge Design Section assists the Field Operations Sectionwhen necessary and will check shop drawings for structures that they have designed.Consultants must check their own shop plans.


Design Section


The specifications and special provisions dictate shop drawing submittal requirements. Thefollowing is a partial list of items that require shop drawings:

♦ Various prestressed concrete beams

♦ Deck panels

♦ Preformed metal deck forms

♦ Structural steel

♦ Segmental prestressed concrete

♦ Retaining wall systems requiring shop drawings by specification

♦ Sound barrier walls

♦ Bearing pads and other structural bearings

♦ Various bridge joints (armor joints, sealed expansion joints, finger joints, etc.)

♦ Bridge protective assemblies

♦ Overhead sign bridges

♦ Concrete piling

♦ Prefabricated pedestrian bridges

Chapter 2 — TxDOT and Bridge DesignSection 3 — Coordination Responsibilities of the

Bridge Design Section


Section 3 Coordination Responsibilities of the Bridge Design Section

Overview

All responsibilities of the Bridge Design Section are actually coordinative because anotherentity is always involved. However, defined coordinative responsibilities are thosefunctions that are not considered a part of the fundamental duties of the Bridge DesignSection but that are important and are carried on daily.

Overweight/Oversize Permits

There are increasing requests for permits to move overweight/oversize loads over statehighways. When requested by the permitting agency, the Bridge Construction Section, withoccasional assistance from the Bridge Design Section or a consultant, makes an assessmentof the effects of these vehicles on the bridges along the proposed route. The BridgeConstruction Section will approve or disapprove accordingly.

Structures Research Management Committee (RMC 5)

TxDOT carries on a very extensive research program, primarily through the HighwayPlanning and Research Program. In the area of Structures and Hydraulics, the BridgeDesign Section and some districts provide technical support for the various projects. Overmany years, this program has generated a significant library of reports on various aspects ofstructural and hydraulic design. Many of the findings have been incorporated in the designspecifications or procedures.

Preliminary Consultation Regarding Structure Type

Preliminary consultation regarding structure type is a very important duty, especially whenbridge plans are to be prepared by the districts or the consulting engineers. Once the PS&Eare submitted it is usually too late to change systems or major details.

Note: For major bridges, especially, interaction with the Bridge Design Section is desirableduring the preparation of preliminary layouts.

Consultation During Plan Preparation by Others

Consultation during plan preparation by others, especially a consulting engineer, canalleviate misunderstandings and avert delays.

Note: No matter how busy the workload, the Bridge Design Section will try to makesomeone available for consultation in a timely manner.




Scheduling Plan Work

Scheduling plan work is usually a negotiation process with the district. The Bridge DesignSection will consider workload and available personnel, allowing the details to be completedto meet a realistic target date. Chapter 5 of the TxDOT Bridge Project Development Manualcontains suggested lead times for submitting bridge layouts.

Oversight

The determination of federal or state oversight for highway projects is discussed inChapter 3 of the TxDOT Bridge Project Development Manual. When the state has oversightresponsibilities, the Bridge Design Section will exercise oversight of all aspects of bridgestructural design.

Interaction with Outside Agencies

Outside agencies are often involved in various aspects of bridge planning and design.Chapter 3, Section 3 of the TxDOT Bridge Project Development Manual lists and describesmany of these agencies. Additionally, the Bridge Design Section has close workingrelationships with the following agencies:

Federal Highway Administration. One of the most important interactions since the adventof the Interstate Highway System has been with the Federal Highway Administration. Theyhave maintained strong bridge sections in Washington (Headquarters), Atlanta (SouthernResource Center), and Austin (Texas Division).

National Committees. The Bridge Design Section represents TxDOT on several nationalcommittees and organizations that furnish information and develop procedures for structuraldesign.

♦ The American Association of State Highway and Transportation Officials (AASHTO)The director of the Bridge Division represents Texas on the prestigious AmericanAssociation of State Highway and Transportation Officials Highway Subcommittee onBridges and Structures. This organization is responsible for writing and revising thestructural design specifications to be followed by all 50 states. They publishconstruction specifications also, but these are modified heavily by our own TxDOTStandard Specifications for Construction of Highways, Streets, and Bridges. Thedirector of the Bridge Division and/or their representatives meet annually with the fullbridge committee. The group is divided into several subcommittees, each with aspecific structural system to monitor for possible improvements to the specification.Specification revisions usually originate in these subcommittees. Texas has been ableto influence the specifications by the results of local research.

♦ The American Railway Engineering and Maintenance-of-Way Association (AREMA)AREMA is the organization that controls everything associated with railwayengineering and maintenance and publishes a specification that must be followed whendesigning structures on or over a railroad.




♦ Transportation Research Board (TRB)The Transportation Research Board is a federal agency that manages transportationresearch projects contracted by universities and other research organizations in the U.S.and Canada.

♦ The American Concrete Institute (ACI)The American Concrete Institute publishes a specification for reinforced concrete that iswidely used for building construction. This is a very important institute, withcontributors from many universities nationwide. AASHTO sometimes draws on theexperience of ACI in revising its specification.

♦ Prestressed Concrete Institute (PCI)The Prestressed Concrete Institute publishes a manual and keeps up-to-date ondevelopments in prestressed concrete.

♦ Post-Tensioning Institute (PTI)The Post-Tensioning Institute publishes a manual and keeps up-to-date ondevelopments in post-tensioned concrete.

♦ The American Institute of Steel Construction (AISC)The American Institute of Steel Construction publishes a specification for structuralsteel design that is widely used in building construction. AISC is active in trying tokeep AASHTO current in steel design and publishes a manual that contains much usefulinformation regarding availability and capability of steel components.

♦ The American Iron and Steel Institute (AISI)The American Iron and Steel Institute is a nonprofit service organization for thefabricated steel industry in the United States and is dedicated to presenting the mostadvanced information available to the technical professions.

♦ The American Segmental Bridge Institute (ASBI)The American Segmental Bridge Institute is a nonprofit organization that provides aforum where owners, designers, constructors, and suppliers can meet to further refinecurrent design, construction, and construction management procedures, and evolve newtechniques that will advance the quality and use of segmental concrete bridges.

♦ The American Society of Testing and Materials (ASTM)The American Society of Testing and Materials develops and publishes specificationsfor all types of materials used in highway construction.

Chapter 2 — TxDOT and Bridge DesignSection 4 — Contractive Responsibilities of the



Section 4 Contractive Responsibilities of the Bridge Design Section

Overview

The Bridge Design Section manages a pool of consultants that expands its capabilities tomeet the demands of large letting volumes. The contracting process is subject to rulesgoverning all TxDOT engineering contracts.


Chapter 3 Design Specifications

Contents:Section 1 — Mandatory Specifications ................................................................................3-2

Section 2 — Guide Specifications ........................................................................................3-4

Section 3 — Industry Recommendations..............................................................................3-6

Chapter 3 — Design Specifications Section 1 — Mandatory Specifications


Section 1 Mandatory Specifications

Overview

There are many specifications available that have a bearing on the design of bridges andother highway structures. This section identifies those specifications that the TexasDepartment of Transportation (TxDOT) considers mandatory for use.

AASHTO Standard Specifications for Highway Bridges

The Standard Specifications for Highway Bridges adopted by the American Association ofState Highway and Transportation Officials (AASHTO) is the most important control overbridge design. It is usually published in full every four years. In the intervening years,Interim Specifications are distributed, which contain the revisions approved on a ballotfollowing the last meeting of the AASHTO Highway Subcommittee on Bridges andStructures. The Federal Highway Administration (FHWA) may, at any time, review designsand details for compliance with these specifications for projects using federal money.

Note: Although these specifications are considered mandatory, a few deviations are madebased on long-time local practice or research.

Copies of these specifications, as well as other AASHTO publications, may be purchasedfrom AASHTO by calling 1-800-231-3475, or at their website athttp://www.transportation.org/.

AASHTO Standard Specifications for Structural Supports for Highway Signs, Luminairesand Traffic Signals (D1.1)

Standard Specifications for Structural Supports for Highway Signs, Luminaires and TrafficSignals is also published by AASHTO. This specification contains a more comprehensivetreatment of wind effects on structures.

ANSI/AASHTO/AWS Bridge Welding Code (D1.5)

The American National Standards Institute (ANSI)/AASHTO/American Welding Society(AWS) Bridge Welding Code combines the recommendations of the three agencies withregard to welding details, methods, and quality tests. Bridge design is no better than thedetails that are generated, and structural steel is particularly detail oriented because offatigue considerations.

Chapter 3 — Design Specifications Section 1 — Mandatory Specifications


AREMA Specifications

The American Railway Engineering and Maintenance-of-Way Association (AREMA)Specifications cover many aspects of railway engineering, including the design of bridgesand culverts that carry railway traffic.

Individual Railroad Company Requirements

Individual railway companies may have their own supplemental requirements, which shouldbe investigated for each project.

Chapter 3 — Design Specifications Section 2 — Guide Specifications


Section 2 Guide Specifications

Overview

Specifications that are not binding to bridge design, but may be useful or even vital to thedesign process are referred to as guide specifications. This section identifies and describesguide specifications recommended by TxDOT.

AASHTO Guide Specifications

AASHTO publishes a number of specifications and manuals, other than those discussed inSection 1, Mandatory Specifications, relating to bridge design that may be useful to thedesigner.

Standard Specifications for Highway Bridges — Load and Resistance Factor Design.This document is a complete rewrite of the bridge specification, with load and resistancefactors based on probability analyses. In coming years, bridges in Texas will be designedusing this specification. This places it in the guidance category.

Guide Specifications for Design and Construction of Segmental Concrete Bridges. Thisdocument contains guidelines for the design and construction of segmental concrete bridges.The guidelines are the recommendations of a team of nationally recognized experts,composed of consulting engineers, contractors, academicians, researchers, state highwayagencies, and federal agency representatives from throughout the United States as well asrepresentatives from Canada, France, Switzerland, and Germany.

The guidelines are comprehensive in nature and embody several new concepts that aresignificant departures from previous design and construction provisions. They areformulated and based on both observed performance of bridges of this type and on recentresearch conducted in the United States and abroad.

This document was originally prepared by the Post-Tensioning Institute under NationalCooperative Highway Research Program (NCHRP) Project 20-7/32 with the title, “Designand Construction Specifications for Segmental Concrete Bridges,” in February 1988. It wassubsequently studied and approved as a guide specification by the AASHTO HighwaySubcommittee on Bridges and Structures in 1989.

Additional AASHTO Publications. There are several other guide specifications, standardspecifications, and manuals published by AASHTO. Some of those include:

♦ Guide Specifications for Alternate Load Factor Design Procedures for Steel BeamBridges Using Braced Compact Sections

♦ Guide Specifications for Fracture Critical Non-Redundant Steel Bridge Members

♦ Guide Specifications for Horizontally Curved Highway Bridges

♦ Guide Specifications for Bridge Railings

Chapter 3 — Design Specifications Section 2 — Guide Specifications


♦ Guide Specifications for Fatigue Design of Steel Bridges

♦ Guide Specifications for Fatigue Evaluation of Existing Steel Bridges

♦ Guide Specifications for Strength Evaluation of Existing Steel and Concrete Bridges

♦ Guide Specification and Commentary for Vessel Collision Design of Highway Bridges

♦ Guide Specifications for Strength Design of Truss Bridges (Load Factor Design)

♦ Guide Specifications for Distribution of Loads for Highway Bridges

♦ Guide Specifications for Structural Design of Sound Barriers

♦ Guide Specifications for Aluminum Highway Bridges

♦ Guide Design Specifications for Bridge Temporary Works

♦ Guide Specifications for Design of Pedestrian Bridges

♦ Standard Specifications for Movable Highway Bridges

♦ Manual for Maintenance Inspection of Bridges

♦ Manual for Condition Evaluation of Bridges

Chapter 3 — Design Specifications Section 3 — Industry Recommendations


Section 3 Industry Recommendations

Overview

Industry recommendations are valuable to the structural designer. The specificationsproduced usually have the backing of leaders in the industry and also experts fromuniversities and other research agencies nationwide.

Concrete

Concrete design is the subject of these publications:

♦ American Concrete Institute Specifications(ACI)

♦ Prestressed Concrete Institute Manual (PCI)

♦ Post-Tensioning Institute Manual (PTI)

Steel

Structural steel design is the subject of these publications:

♦ American Institute of Steel Construction Specification and Manual (AISC)

♦ American Iron and Steel Institute Manual (AISI)

Materials

Specifications for a wide range of materials are contained in several volumes of AmericanSociety of Testing and Materials publications.


Chapter 4 Geometric Restraints

Contents:Section 1 — Bridge Width....................................................................................................4-2

Section 2 — Bridge Span Length .........................................................................................4-4

Section 3 — Horizontal and Vertical Clearances .................................................................4-5

Section 4 — Alignment.........................................................................................................4-8

Section 5 — Stage Construction .........................................................................................4-13

Chapter 4 — Geometric Restraints Section 1 — Bridge Width


Section 1 Bridge Width

Overview

Bridge width depends solely on the width of the highway except in unusual cases.Geometrically, bridges are just a small part of the highway. From the driver’s standpoint,bridges need to blend inconspicuously into the perception of the road. This need makesbridge widths, alignment, and clearances subject to the requirements of the highwayengineer.

Refer to the Texas Department of Transportation Department (TxDOT)Roadway Design Manual for guidance on highway design.

Background

Roadway widths covered by Bridge Design Standards have ranged from 16 ft. to 48 ft. andfrom 7.2 m to 13.2 m during the past 80 years. For several years, bridges had curbs located2 ft. outside of the traffic lane. Texas began providing graveled shoulders on majorhighways earlier than most states. With continuing increases of highway speed, it becameevident that many vehicular accidents happened at the beginning of bridges where thehorizontal clearance became restricted. Texas began a campaign for shoulder width bridgesin the 1950s, but the recommendation did not appear in the American Association of StateHighway and Transportation Officials (AASHTO) Specifications until 1969 because of theconsiderable cost of the additional bridge width.

Current Status

Today, virtually all bridges in Texas are as wide as the approach roadway, includingshoulders. Curbs are not used except to protect a pedestrian walkway on a low-speedhighway. The nominal face of bridge railing is located at the outside edge of the shoulder.Traffic lanes are usually 12 ft. (3.6 m) wide but may be reduced for low-volume orextremely crowded conditions and may be increased for sight distance around a horizontalcurve. Shoulders vary from 2 ft. (0.6 m) to 10 ft. (3.0 m) depending on traffic volume andstructure function. This results in a large number of bridge widths. The Bridge Divisionprovides standard bridge details for a few of the most repetitive widths that are less likely tohave complicated geometry.

Design Practice

On divided highways, separate bridges are used for each direction of traffic, unless themedian is less than 30 ft. (9.0 m) wide between inside lane edges. In this case, bridgesurfaces are usually flush with the pavement across the median. A concrete traffic barrier isusually constructed at the center of the median. The resulting bridge can be quite wide whenthere are multiple lanes in each direction. To ensure against transverse expansion andcontraction problems, a longitudinal open joint is recommended when the bridge width

Chapter 4 — Geometric Restraints Section 1 — Bridge Width


exceeds 120 ft. (36.0 m). For simplicity of design, the joint should be at the mediancenterline. On long span structures, differential deflection can cause cracking in theconcrete traffic barrier, which might justify moving the joint to one side of the barrier.

Except for unusual situations, the overall width of bridge decks is 2 ft. (0.6 m) more than thedistance between the nominal faces of outside railing. Most standard bridge railings occupyless than 1 ft. (0.3 m) of deck width. The safety-shape or straight-sided traffic railings areslightly wider in their lower part, but this is not considered sufficient encroachment to affectoperation of the shoulder. The nominal face of railing is set at 1 ft. (0.3 m) from the deckedge to allow the use of different railing with the same standard details. This dimension isalso recommended for all non-standard bridges. Bridge widths for the current BridgeDesign Standards are shown in Figure 4.1.

Figure 4-1. Standard Bridge Widths (Online users can click here to view this illustration inPDF.)

Chapter 4 — Geometric Restraints Section 2 — Bridge Span Length


Section 2 Bridge Span Length

Overview

Bridge length depends on terrain, hydraulics, prescribed clearances, or aesthetics, and iscontrolled by economics and the capability of available structural systems.

Grade Separations

Prescribed clearances and header slopes govern highway grade separation span lengths.Sometimes aesthetic considerations may dictate longer-than-necessary spans to give theseparation a more open look underneath. The larger prestressed concrete beams havebecome so economical that they are used where shorter spans would suffice.

Stream Crossings

Stream crossing span lengths usually have a main span that straddles the stream. For smallstreams, where hydraulic considerations are minimal, the bridge is often divided into equalspans for ease of construction. However, no bents or piers should be placed in erodablestreambeds. River crossings almost always have a main span across the center of flood flow;the purpose being to discourage an accumulation of drift on the piers. There are no warrantsfor determining the length of this span; it is a matter of engineering judgment.

Marine Bridges

Intracoastal canal and international shipping lane main spans are subject to U.S. CoastGuard regulations. Approach spans are determined by economics and/or aesthetics.Navigable rivers as determined by the Federal Highway Administration (FHWA) are alsosubject to U.S. Coast Guard Regulations.

Chapter 4 — Geometric Restraints Section 3 — Horizontal and Vertical Clearances


Section 3 Horizontal and Vertical Clearances

Overview

Clearances are established by AASHTO, American Railway Engineering and Maintenance-of-Way Association (AREMA), FHWA, U.S. Army Corps of Engineers, U.S. Coast Guardand, to some extent, local authorities. Minimum horizontal and vertical clearances forhighway bridges are established in the TxDOT Roadway Design Manual, repeated in theTxDOT Bridge Project Development Manual, and tabulated here for ready reference.Clearances to railroads are specified by AREMA. Intracoastal canal clearances aredetermined by the U.S. Army Corps of Engineers. International shipping lane clearancesmust be negotiated with the U.S. Coast Guard.

Highway Grade Separations

Clearance measurements for highway grade separations are depicted in Figure 4-2. For acomplete listing of horizontal and vertical clearances for specific highway functionalclassifications, see the Roadway Design Manual.

Highway Grade Separation ClearancesVertical 14'-6" (4.42 m) Absolute minimum

16'-6" (5.03 m) To be provided over all roadways if possibleand mandatory for new construction overinterstate highways

Horizontal 1'-6" (0.46 m) Absolute minimum from face of curb orbarrier

10'-0" (3.05 m) From edge of travel lane on low-speed, low-volume roadways

16'-0" (4.88 m) From edge of travel lane on medium-volumeroadways and freeway ramps

30'-0" (9.14 m) From edge of travel lane on high-volumeroadways and all freeway main lanes

Note: Special conditions that would severely increase structure cost may justify negotiation of theseclearances with the TxDOT Bridge Division.

Stream Crossings

Stream Crossing ClearancesVertical 2'-0" (0.61 m) Desired, above design high water

1'-0" (0.30 m) Absolute minimum, above design high waterHorizontal As determined by topography and hydraulics



Railroad Overpasses

See Figure 4-2.

Railroad Overpass ClearancesVertical 23'-0" (7.01 m) Absolute minimum

24'-3" (7.39 m)to

26'-0" (7.92 m)

May be required for electric powered trains

Horizontal 9'-0" (2.59 m) Absolute minimum (12' crash walls required)12'-0" (3.66 m) Desirable minimum (6' crash walls required)25'-0" (7.62 m) Minimum to eliminate crash walls

Railroad Underpasses

Railroad Underpass ClearancesVertical Same as for Highway Grade Separation Structures

Horizontal Same as for Highway Grade Separation StructuresNote: Although the horizontal clearance criteria for a railroad underpass are the same as for a highway gradeseparation structure, consideration should be given to using appropriate barrier railing through the railroadunderpass, allowing for a reduced horizontal clearance. Given the high cost of this structure type, thispractice will greatly reduce bridge cost by reducing the required span length.

Pedestrian Bridges and Non-redundant Bridge Supports

Pedestrian Bridge and Non-redundant Bridge Support ClearancesVertical 17'-6" (5.33 m) Minimum

Horizontal Desirably greater than highway grade separation structures

Sign Bridges

Sign Bridge ClearancesVertical 17'-6" (5.33 m) Minimum

Horizontal 2'-0" (0.61 m) Absolute minimum from face ofbarrier railing

Note:: Additional vertical clearance may be required by some area offices based on experience.

Intracoastal Canal Bridges



Intracoastal Canal Bridge ClearancesVertical 73'-0" (22.25 m) Above mean high water

Horizontal 125'-0" (38.10 m) From center of channel

Bridges over International Shipping Lanes

Clearances are subject to negotiation with the U.S. Coast Guard.

Figure 4-2. Clearance Measurements (see following explanatory notes) (Online users canclick here to view this illustration in PDF.)

Explanatory Notes for Figure 4-21. Horizontal clearance from edge of curbed roadway to obstruction. Refer to the TxDOT

Roadway Design Manual for specific criteria of roadway functional classification.

2. Horizontal clearance from outside edge of exterior lane (uncurbed) to obstruction.Refer to the TxDOT Roadway Design Manual for specific criteria of roadwayfunctional classification.

3. Horizontal clearance from the centerline of tracks to obstruction. Use 25'-0" or greaterto avoid the need for crash walls. Minimum horizontal clearance of 9'-0" with crashwalls that are 12'-0" above track elevation. Minimum horizontal clearance of 12'-0"with crash walls that are 6'-0" above track elevation.

Chapter 4 — Geometric Restraints Section 4 — Alignment


Section 4 Alignment

Overview

The subject of alignment covers horizontal and vertical curvature of the profile and/orstation line and the cross-slope of the deck surface.

Background

In the early days, highway engineers were satisfied with bridges that were straight, square,and relatively flat. It gradually became evident that bridges could handle other types ofalignment, although with considerably more complexity in the details. Presently, curves,skews, variable widths, and crown rollouts are normal. About the only alignment that is notcompatible with bridges is the spiral curve. Spirals are still used occasionally for highwayalignment, but they are usually approximated by three centered circles for use in bridgeframing.

Current Practice

Highway alignment follows the guidelines given in the TxDOT Roadway Design Manual.Bridge alignment conforms to these alignments and is usually a “given” on the preliminarylayouts.

Design Recommendations

Horizontal Curvature. Horizontal curvature up to 5 degrees on wide bridges and 10 degreeson narrow connection structures can be expected. Curves up to 20 degrees haveoccasionally been used on “button hook” ramps and turnarounds. Horizontal curvature ofbeams, with the exception of pan form girders, can be handled gracefully in cast-in-placestructures, but these have not been economical in Texas for many years. The preferredsupport system is precast prestressed beams. Since the beams must be straight, overhangwidth to the curved deck edge may limit the span length. Figure 4-3 shows this relationship.If the curvature/span length combination exceeds the capability of the deck slab, the spanmust be decreased or other measures must be considered, such as the use of curved steelgirders.

Vertical Curvature. Extreme vertical grade can cause construction problems but seldominfluences structure type. Grades over 5 percent call for extra care during concreteplacement. The concrete tends to flow downhill during finishing operations. Thicker slabspans are more sensitive than deck slabs. Elastometric bearing details for prestressedconcrete beams require special consideration for grades over 5 percent, Extreme verticalcurvature can seriously affect forming methods for deck slabs on prestressed beams,especially if precast concrete deck panels are used. Crest curves cause extra deck depth inthe middle of the span. Sag curves cause extra depth at the ends. If the deck slab is cast onremovable forms, this extra depth can be accommodated in the haunch depth over each



beam. Even so, it is necessary to set the haunch depth carefully to avoid constructionproblems.

Based on AASHTO SlabGrade 60 Reinforcing

*Adjust to 1/4 pt. of Flanges for Steel Beams

Figure 4-3. Guidelines for Horizontal Curvature Using Prestressed Beams (Online userscan click here to view this illustration in PDF.)

Gradeline. Current practice is to set haunch depths that will keep the top of beam at orbelow the bottom of the slab. Extra reinforcing is required if the haunch depth exceeds3.0 in. (75 mm). If precast concrete deck panels are used, the problem is more critical.Special grading details may be required to accommodate tall haunches. Bearing seatelevations may require lowering for sag curves. Variable camber in prestressed beamsaggravates the problem. It may not be possible to cover all these variations in the designstage. Contractors have become accustomed to adjusting the gradeline after takingelevations on the tops of the erected panels.

Cross-slope or crown for bridges is 1 percent minimum, 2 percent desirable. If the structureis more than two lanes wide, the outer lanes are usually sloped 2.5 percent to facilitatedrainage. Cross-slopes can transition into superelevations as much as 8 percent on curvedstructures. Superelevation above 5 percent can cause problems with concrete placement thesame as steep grades. If such deck slopes cannot be avoided, the construction engineersshould be alerted to the possible need for special concrete placement requirements.



Superelevation also affects clearances between deck slab and beam, especially when precastconcrete panels are used. Superelevation creates an apparent sag vertical curve along theprestressed beam, which is a chord to the curvature.

When vertical curvature and superelevation exist, and the effect of beam camber is added,drastic measures may be required, especially with sag curves and panel deck construction.All of this can be accommodated, but extreme caution should be exercised and detailedgeometric computations made.

Roadway Design System. Deck dimensions, beam framing, bearing seat elevations, webcutting, and bent locations must be accurately calculated to fit the prescribed alignment.This calculation is the responsibility of the bridge designers. The Roadway Design System(RDS) is a geometric computer program, originally developed in Texas and formerly usednationwide. The program has several bridge oriented capabilities for slabs, beams, andgirders and is used exclusively by the Bridge Design Section for prestressed concrete beamspans on curves. The more important bridge routines in RDS are the following:

♦ SLAB. Computes and tabulates edge dimensions and areas for deck slabs of allconfigurations. Slab edges can be plotted.

♦ SLEL. Will produce a tabulation of distances, surface elevations, bottom of slabelevations, and bottom of slab plus dead load deflection along the boundaries of theslab.

♦ FOPT. Computes and tabulates framing dimensions for beam spans or continuousgirders according to one of several programmed options. Framing diagrams can beplotted.

♦ BMGD. Will produce a tabulation of surface elevations, bottom of slab elevations, andbottom of slab elevations plus dead load deflection along the centerline of each beam.

♦ VCLR. Computes vertical distance from a roadway surface to chorded beam lines. It isused to calculate vertical clearances and to check beam haunch within a span.

Contour plotting is also available in the program. Refer to the Roadway Design SystemManual for further details.

There is a company that maintains the RDS program for national use. However, theInformation Systems Division of TxDOT has performed most of the maintenance of theprogram’s bridge commands in the past few years. The program has also been madecompatible with metric dimensions. Consultants should be careful to use the most recentversion of the program.

Other Software. IGRDS, a computer software roadway design system, is anAASHTOWARE product available from AASHTO.

A new geometry program, called GEOPAK, is being used in most highway applications. Itis very useful in performing the usual highway plan functions, though the company has notyet been able to provide good bridge routines. If there are significant bridges in a project,the alignment file must be duplicated in the RDS format for use in bridge framing.



One problem with RDS is that the roadway surface must be defined by radial cross-slopesfrom only one profile grade line. In varying roadway widths, where ramps are convergingor diverging, it may be necessary to adjust cross-slopes at close intervals along the mainprofile grade line to provide a smooth transition to the ramp grade and avoid edge profileproblems. Contours can be used to advantage in this situation.

Superelevation Transition. Superelevation transition across a varying width roadway cancause unsightly lines on the outside railing. This usually occurs where ramps enter or leavethe main structure. Relative grades between the two also have an influence. Highwayengineers are better able to work out this problem, but it appears often to be overlooked orloosely handled. It is recommended that bridge engineers consider this situation carefullybefore setting cross-slopes for framing computations. Contour plots and a plot option ofSLEL can be useful in these considerations.

Under certain conditions, a combination of superelevation transition and vertical curvaturewith a constant roadway width can cause sags or humps on the outside of the bridge. Bothare unsightly, and sags can pond water on the roadway surface. This problem is usuallycorrected by highway engineers, but it would be advisable for the bridge designers to verifythe outside lines by contours or pavement edge profile plots.

Superelevation transitions can have an adverse effect on beam haunch. This effect can beminimized by starting and ending a transition at a bent. Figure 4-4 shows the built-in andrecommended optional methods for handling superelevation transition in RDS.



Figure 4-4. Superelevation Transition According to Roadway Design System (Online userscan click here to view this illustration in PDF.)

Chapter 4 — Geometric Restraints Section 5 — Stage Construction


Section 5 Stage Construction

Overview

Stage construction is required when traffic must be diverted onto a portion of an existingbridge while part of the new structure is built, then moved over for reconstruction of the firstpart. This section is provided to give the bridge planner/designer some guidelines thatgenerally apply for all staged construction. Topics include existing structure removal, newsubstructure, new superstructure, and temporary railing.

Existing Structure Removal

Texas Standard Specifications, Item 496 “Removing Old Structures,” outlines requirementsfor the removal of existing structure.

The partial removal of the existing structure begins with the cutting and removal of the slab.The location of the cut is called the breakback. The approximate location of the breakbackis determined through coordination with the traffic and highway engineer and is based onlane width requirements of both the new structure and the partial structure to remain inplace. The exact breakback point should be determined by the bridge designer and is basedon the structural capacity of the existing structure.

The breakback is generally located over a beam and must be supported by a stablesubstructure. After the slab is cut and removed, the beams are removed and thesubstructure, or a portion thereof, is demolished. If necessary, footings are removed anddrilled shafts and piles are cut and removed to a distance a minimum of 2 ft., or as specifiedin the plans, below the proposed ground.

New Substructure

The following are guidelines for the design of the new substructure.

Foundations. Consideration must be given to the room required for drilled shaft and pilinginstallation. Both drilled shafts and piling require a 1 ft. minimum horizontal clearance fromedge of foundation to the obstruction. Ideally, there should be no vertical obstruction aboveeither type of foundation. Special drilled shaft rigs are now available that can work with aslittle as 6 ft. of headroom. This equipment is quite expensive, and placement of reinforcingsteel and concrete is very difficult. Contact the TxDOT Bridge Division - GeotechnicalBranch for information on the practicality and cost of these types of shafts.

The only way to install piling in limited headroom is to drive and splice short sections ofsteel piling. This process is seldom practical or cost effective, and should be avoided.

If possible, avoid the location of the existing foundations that remain. For widenings,foundations should be of similar type as those remaining in use.

Chapter 4 — Geometric Restraints Section 5 — Stage Construction


Abutments. At the stage construction joint, it is difficult to leave reinforcing steel projectingfrom the abutments for splicing because of the conflicts with the temporary shoring thatmust retain the fill. Instead, locate foundations (drilled shafts or piling) close to the stageconstruction joint and dowel the two sides of the cap together, or provide a sealed expansionjoint.

Interior Bents. If possible, use independent bents. If a single structure is required, thereinforcing steel can be spliced together using a lap or mechanically coupled together. Ifsplicing is used, adequate horizontal and vertical clearances must be provided to account forthe projecting reinforcement. The exposed reinforcement must be protected. If availableclearances are limited, use mechanical couplers or butt welds. Due to the complexity ofcouplers and welds, accurate details and proper structural detail notes are essential.

New Superstructure

The following are guidelines for the design of the new superstructure.

The location of the stage construction joint in the slab and the available clear distance forsplicing the mat reinforcing are critical factors in the slab design. The stage constructionjoint can be placed over a supporting beam or in a bay between beams. However, placingthe stage construction joint over a supporting beam is the preferred method. When placingthe joint over a supporting prestressed beam, the joint must be located 2 in. beyond thecenterline of the beam to grab the R-bars with the first pour. Prestressed concrete panels aretypically not allowed in the second placement in the bay adjacent to the construction joint.When placing the joint between beams, locate the joint at the quarter point of the beamspacing.

Joints should be located so that space for minimum reinforcing steel laps and 1 in. of coverbeyond the ends of the bars is provided. The available construction clear distance may limitthe available length required for an adequate lap length. If the clear distance is inadequate,mechanical couplers can be utilized. However, there are concerns about the performance ofa construction joint using couplers in both mats, particularly in salt areas. If couplers areused, be sure the appropriate specifications are supplied. Consideration of raising the gradea few inches to allow the top mat to be lapped should be given. Shorter laps might bejustified based on the AASHTO provision (As required /As provided) in areas where theslab has excess capacity.


Chapter 5 Preliminary Considerations

Contents:Section 1 — Materials...........................................................................................................5-2

Section 2 — Structure Type..................................................................................................5-9

Section 3 — Economics......................................................................................................5-26

Section 4 — Aesthetics .......................................................................................................5-27

Section 5 — Corrosion Problems........................................................................................5-28

Section 6 — Railing............................................................................................................5-29

Chapter 5 — Preliminary Considerations Section 1 — Materials


Section 1 Materials

Overview

Availability of materials is generally not a factor in determining the most suitable type ofstructure for a given location. Concrete and steel are the basic ingredients of most structures,and they are available to every county in the state. While bridges are primarily concrete andsteel, aluminum is used very sparingly in railing and pipe; plastics are used for smalldiameter pipe; asphalt is used for overlays; neoprene is used for bearings; and butyl rubber isused for railroad underpass waterproofing.

Except for reinforcing steel, only brief descriptions are given here.

Concrete

Concrete is described by class, which identifies its strength, cement content, water/cementratio, and coarse aggregate type according to the item “Portland Cement Concrete” of theTexas Department of Transportation (TxDOT) Standard Specifications for Construction ofHighways, Streets, and Bridges. Concrete may be made from many different sources ofcement, fine and coarse aggregate, and water, but all materials must meet the requirementsof the specification.

Various additives are allowed or required for certain conditions of use. The use of fly ash toaugment or replace some of the cement is gaining acceptance. Silica fume has also beenused. It has been demonstrated that high strength concrete, around 13,000 psi compressivestrength, can be produced from Texas aggregates and successfully placed in the forms forcertain bridge members. These ingredients are also being used to develop a “highperformance concrete” (HPC) with emphasis on density to provide better resistance tochloride attack.

Reinforced concrete is the term applied to concrete containing reinforcing bars designed toresist any tension that may occur in the member. Virtually all bridges contain somereinforced concrete. The concrete is usually mixed nearby and trucked to the job site.

Prestressed concrete is the term applied to high-strength concrete containing very high-strength steel that has been stretched and anchored to the concrete with sufficient force tosignificantly reduce tension from occurring in the member. When the concrete is placedbefore the steel is stretched, the member is said to be “post-tensioned.” When the steel isstretched before the concrete is placed, the member is said to be “pretensioned.” Post-tensioned structures are used sparingly, but pretensioned precast concrete beams are themainstay of Texas bridge construction. Type IV beams up to 135 ft. long are possible tomanufacture and transport. Greater lengths are possible with the use of high performanceconcrete. Type VI (MOD) beams can be used for spans up to 175 ft. in specialcircumstances. Use of long beams, however, depends on the accessibility of the bridge siteby transporting trucks.



Structural Steel

Structural steel is available in many shapes and sizes. Much of the structural steel ismanufactured elsewhere, but fabrication is usually performed in or near the state. Furtherdiscussion of structural steel can be found in Chapter 7 of this manual.

Prestressing Steel

Prestressed steel is a very high-strength material, which is discussed further in of thismanual.

Reinforcing Steel

Background. There have been many changes in the strength and configuration ofreinforcing bars in the history of TxDOT. Smooth bars cold twisted to improve bond wereused early, but soon outlawed by the specifications. All bars were square for awhile and,even into the late 1940s, #9, #10, and #11 bars were square. Oil well sucker rods were usedoccasionally during World War II because of a scarcity of regular reinforcing bars. For non-specification work it was possible to find anything from barbed wire to old car partsreinforcing the concrete.

Variations through the years in the specification requirements for reinforcing bars are shownin tables for years 1918-1953 and 1953-1988.

Ductile structural grade steel was used until the early 1950s. Rail steel was added, only tobe removed in the late 1970s and added back in the 1980s. Deformations were a bigconcern of the 1940s but the questions were put to rest by ASTM 305-47T. The early 1960ssaw the availability of #14 and #18 bars established. The 1973 American Association ofState Highway and Transportation Officials (AASHTO) specification ushered in high-strength reinforcing steel and put a limit on stress range to avoid fatigue problems.Weldable reinforcing steel was covered by ASTM A706. Grade 75 bars were considered forconcrete but abandoned because of the absence of a yield plateau in the stress/straindiagram. Grade 75, size #18S bars were used for anchor bolts by one light polemanufacturer. Epoxy coating of reinforcing bars was introduced in the late 1970s.

Design of reinforcing steel requires analysis of the complex interaction with concrete slabs,beams, columns, and footings. For service load design, an allowable stress is specified thataccounts for a reasonable factor of safety. Load factor design, which has become thestandard method, allows the reinforcing steel to reach yield under the action of loadsfactored up to provide safety. With load factor design, service load stresses must usually becalculated to insure that crack width and fatigue stress limits are not exceeded.



Chronology of Reinforcing Steel Specifications (1918-1953)AASHTO

SpecificationMaterial Specification Special Requirements fs allowable

(ksi)1918

(T.H.D.)Yield ≥ 33 ksiBend 180° over one diameter pin

O.H., Mild, or MediumPlain, Twist, and Deform

16

1926(T.H.D.)

A15-14 O.H., Deform, or Plain Twistwith engineer’s approval

16

1931 A15-30 O.H., Str. OnlyNo Twist

LL 16DL 24

1935 A15-33 (Mod.) O.H., Str. OnlyNo Twist

16

1941 A15 O.H., Str. or Int., Deformapproved by engineerNo Twist

18

1944 M31-42 O.H., Str. or Int., Deformapproved by engineerNo Twist

Str. 18Int. 20

1949 M31-38 (Deform) O.H., Str. or Int.,All DeformNo Twist

Str. 18Int. 20

1953 M31-52 (Billet)M42-48 (Rail)A305-50T (Deform)

O.H.All DeformNo Twist

Str. 18Int. & Hard 20

T.H.D. = Texas Highway DepartmentA15, etc. = ASTM Spec.M31, etc. = AASHTO Spec.O.H. = Open hearthTwist = Cold twisted plain barsDeform = Deformations to improve bond

Str. = Structural gradeInt. = Intermediate gradeHard = Hard gradeMild = Structural gradeMed. = Intermediate grade



Chronology of Reinforcing Steel Specifications (1957-1988 Interim)AASHTO

SpecificationMaterial Specification Special Requirements fs allowable

(ksi)1957 A15-54T (Billet)

A16-54T (Rail)A305-53T (Deform)



1961 A15 (Billet)A16 (Rail)A408 (#14 & #18)A305/408 (Deform)



1965 A15 (Billet)A16 (Rail)A408 (#14 & #18)A305/408 (Deform)

O.H., E.F., B.O.No Twist


1969 A615 (Billet Gr. 40)A42 (Rail Gr. 50)


20

1973 A615 (Billet Gr. 40)A615 (Gr. 60)A42 (Rail Gr. 50)


Gr. 40/20Gr. 60/24

1977 M31 (Billet Only)A706 (Weldable)

Special Bend RequirementsFor #14 & #18 Gr. 50

Gr. 40/20Gr. 60/24

1983 A615 (Billet Only)A706 (Weldable)

Special Bend Requirementsfor #14 & #18 Gr. 50

Gr. 40/20Gr. 60/24

1988Interim

A615 (Billet)A616 (Rail)A706 (Weldable)

Supplementary RequirementsS1 for Bending A616

Gr. 40/20Gr. 60/24

A15, etc. = ASTM Spec.M31, etc. = AASHTO Spec.O.H. = Open hearthE.F. = Electric furnaceB.O. = Basic oxygenTwist = Cold twisted plain barsDeform = Deformations to improve bond

Str. = Structural gradeInt. = Intermediate gradeHard = Hard gradeGr. 40 = 40 ksi yield stressGr. 60 = 60 ksi yield stress

Probably the most volatile part of the design specification is tensile lap splice lengths. Foryears, bars were lapped enough to develop the allowable working stress in bond. Allowablebond stresses changed often and were finally eliminated from the specification. Otherfactors began to be studied in the late 1960s. The studies continued until the early 1980s.1The design specification in 1977 may have over reacted to the first research and splicelengths became long and variable. The later research reexamined previous findings anddeclared the problem less severe. This research was taken into account for the 1989American Concrete Institute (ACI) Code, but splice lengths still came out long and variable.Current AASHTO Specifications are similar to the ACI Code. Refer to the TxDOTBridge Detailing Manual for tables of development lengths and lap splices.



Welded splices have been used under some conditions. Bars #7 and smaller are lapped 4 in.and welded both sides. Larger bars are butt welded. Controlled chemistry is usuallyrequired for welds. Weldable ASTM A706 reinforcing bars are now readily available.

Originally, mechanical splices were usually confined to bars in compression. Metal filledsleeve splices became acceptable for tension splices. Adjacent bridge decks on a large stageconstruction project used dowels in tapered thread couplers, after an extensive test programto verify their fatigue performance. More manufacturers developed mechanical splicesystems, so that there are now several acceptable alternatives for tension splices.

Fabrication problems have occurred with bending of the harder steels, especially rail steel.Bars have been known to break when thrown off the truck. This problem was supposedlycorrected by Supplementary Requirement S1 to ASTM A616. Bending tolerance hassometimes been a problem, especially with truss bars formerly used in deck slabs.

The truss bar problem was usually not discovered until construction was in progress. Trussbars on curved decks with chorded beams invariably gave trouble, so truss bars are nolonger recommended for any type of deck. Failure to adequately tie a mat or cagesometimes resulted in movement of bars due to foot traffic or concrete placement.Congested reinforcement has been a frequent complaint in columns, caps, and prestressedconcrete members.

Maintenance problems have been caused by design errors, mislocation of splices, floatingcages, lack of cover, and rust. There is currently no acceptable concrete reinforcing materialthat will not rust in the presence of moisture and air. Epoxy coating has been developed toslow down the corrosion process, but it is sometimes difficult to get properly coated barsfrom the factory into the bridge. Lap splice lengths must be increased as much as 50 percentfor epoxy coated bars.

The design guidelines for reinforcing steel are covered here because it is common to moststructural components constructed on Texas highways.

Current Status. For bridge members, all reinforcing steel should be Grade 60 (420 Mpa)except that longitudinal bars in drilled shafts may be designed as Grade 40 (300 Mpa) toreduce required lap and development lengths. This steel, as well as spirals for columns anddrilled shafts, prestressed concrete beam dowels, and weldable reinforcing bars are allsubject to the requirements of Item 440 of the TxDOT Specification for Construction ofHighways, Streets, and Bridges, which refers to all appropriate American Society of Testingand Materials (ASTM) specifications.

Mechanical splices are allowed or required by plan note. Requirements for mechanicalcouplers are covered by special provision to Texas Standard Specification, Item 440“Reinforcing Steel” and by Departmental Specification DMS-4510.

Design Recommendations. Reinforcing steel design is well covered in the AASHTOSpecifications and the sections that follow. A few observations are made here for emphasis.

♦ The modulus of elasticity of all reinforcing steel should be taken as 29,000 ksi.

♦ In direct load and flexural calculations, linear variation of strain is assumed.



♦ In service load design, stress in the steel is the modulus of elasticity times the calculatedstrain.

♦ In load factor design, stress in the steel is the modulus of elasticity times the calculatedstrain except that, where strain exceeds 0.00207, the stress is at yield, 60 ksi.

♦ Spiral reinforcing steel used for ties in columns should be #3 at 6 in. pitch for columns30 in. round or less and #4 at 9 in. pitch for all others.

♦ Longitudinal reinforcing steel in new columns should have an area of at least 1 percentof the gross concrete area, regardless of the relationship between actual and requiredcapacity.

♦ #18 bars should only be used when the project will require more than 40,000 pounds ofthese bars. Bending is discouraged and lap splicing is prohibited.

♦ Hooks and bends must always conform to the requirements of the AASHTOSpecification to prevent excessive breakage of Grade 60 (420) bars.

♦ Tension lap splice lengths shall conform to the requirements of the current AASHTOspecifications for reinforcing steel.

♦ In consideration of possible coarse aggregate size and to avoid congestion, thefollowing minimum clear spacings should be observed:• Superstructure — 2.5 in.• Substructure — 3.0 in.• Precast Members — 1.5 in.• #18 Bars — 3.5 in.

♦ Avoid splicing in an area where reinforcing steel from an intersecting member exists.Congestion is bound to occur because of spacing tolerances. Splicing cap bars over acolumn should be avoided.

♦ Designed compression splices in flexural members are not recommended.

♦ Splicing in regions of maximum stress is not recommended, but is permissible.

♦ Mechanical splices should be allowed, as an alternate to lap splices, in large epoxycoated bars and should be required in all bars where clearance is doubtful. If possible,welded splices should be allowed as an alternate to mechanical splices.

♦ Requirements for weldable steel and mechanical couplers are covered by TexasStandard Specifications Item 440 “Reinforcing Steel” and the special provisions thereto.

Timber

Fender systems and ferry landings use timber because of its resilience. Recently timber wasused in the rehabilitation of the approach spans to an off-system suspension span. Timberpilings, used for many years in Texas bridges, are no longer recommended.



Aluminum

Aluminum has yet to become useful for structural members in bridges. Early aluminumbridge railing was mostly ornamental because it had very little impact strength. With theadvent of the present railing design specification, the aluminum industry developed an alloythat had sufficient toughness to be cast into rail posts. This type of railing was usedextensively for several years and is still available as a standard. Extruded semi-elliptical railmembers are fastened to the cast posts.

Neoprene

Neoprene, which is a polychloroprene polymer originally patented by DuPont, has beenused for bridge bearings since the advent of prestressed concrete beams in the middle of the1950s.

Fiber Reinforced Polymer (FRP) Composites

Glass and carbon fiber reinforced composites are being studied for use in highwaystructures. Some experimental projects using both structural shapes and concrete reinforcingbars have been developed.

Miscellaneous Materials

Many other materials have been used in conjunction with primary construction materials.Some of these are epoxy, polyester, Teflon, rubber fabric, linseed oil, silane, siloxane,asphalt, coal tar, and concrete additives such as fly ash and silica flume.

Chapter 5 — Preliminary Considerations Section 2 — Structure Type


Section 2 Structure Type

Superstructure

Selecting an appropriate superstructure type is a critical factor in the planning and designprocess. District design engineers, area engineers, or their consultants usually make thechoice of superstructure type, as they prepare the preliminary bridge layouts.

The following figures illustrate the most common superstructure types currently used byTxDOT:

♦ Figure 5-1: Cast-In-Place, Simple Slab Span

♦ Figure 5-2: Cast-In-Place, Continuous Slab Unit

♦ Figure 5-3: Cast-In-Place, Pan Form Spans

♦ Figure 5-4: Prestressed - Precast, I-Beam Slab Span

♦ Figure 5-5: Prestressed - Precast, Double Tee Beam Spans

♦ Figure 5-6: Prestressed - Precast, TxDOT Box Beam Spans

♦ Figure 5-7: Prestressed - Precast, Slab Beam Spans

♦ Figure 5-8: Prestressed - Continuous Slab Units

♦ Figure 5-9: Prestressed - Precast, U-Beam Spans

♦ Figure 5-10: Segmental, Continuous Box Girder Units

♦ Figure 5-11: Segmental, Simple Span Box Girder Units

♦ Figure 5-12: Steel, I-Beam Units

♦ Figure 5-13: Steel, Continuous Plate Girder Units

♦ Figure 5-14: Steel, Continuous Trapezoidal Girder Units

The figures show the economical and practical span limits, and some advantages anddisadvantages of each superstructure type.

In some cases the district may have a preference for certain structure types. If there is anydoubt as to the proper design for the situation, district personnel should contact the bridgeproject development manager or the director of Bridge Design for assistance in determiningstructure type and span lengths.

Currently, over 30,000 bridges and bridge class culverts exist on the state highway system.The most common types of superstructure in use, and a quantity breakdown, are shown in"Types of Superstructure and Quantity Breakdown of On-System Bridges in Texas (FY2000)" table to further aid in selecting a superstructure type.



Figure 5-1. Cast-in-Place, Simple Slab Span (see following explanatory notes) (Onlineusers can click here to view this illustration in PDF.)

Explanatory notes for Figure 5-1

Approximate Superstructure Depth (ft)Skew

Span (ft) 0°°°° 15°°°° 30°°°° 45°°°° 60°°°°20 1.083 1.083

25 to 26.250 1.333 1.333 1.120 1.04030 1.458 1.458 1.250 1.120 0.96035 1.460 1.250 1.04040 1.460 1.080

Advantages Disadvantages1. Minimum depth for short spans2. Ease of design and detail3. Aesthetic for small stream crossings

1. Deck joints at each bent2. Complicated for skews over 15 degrees3. Not the most economical solution4. Requires formwork support5. Limited span length



Figure 5-2. Cast-in-Place, Continuous Slab Unit (see following explanatory notes) (Onlineusers can click here to view this illustration in PDF.)

Explanatory Notes for Figure 5-2

Approximate Superstructure Depth (Ft)Skew

Span (ft) 0°°°° 30°°°°20 1.00 1.0025 1.12 1.1230 1.33 1.3335 1.50 1.50

Advantages Disadvantages1. Absolute minimum depth2. No deck joints3. Aesthetic for small stream crossings

1. Limited to 30 degree skew2. Not the most economical solution3. Limited span length4. Requires formwork support



Figure 5-3. Cast-in-Place, Pan Form Spans (see following explanatory notes) (Online userscan click here to view this illustration in PDF.)


Approximate Superstructure Depth (ft)Skew

Span (ft) 0°°°° 14°°°°02'00" 26°°°°34'00" 36°°°°52'00" 45°°°°30 to 34.167 2.00 2.00 2.00 2.00 2.0040 to 40.833 2.75 2.75 2.75 2.75 2.75

Advantages Disadvantages1. Absolute minimum cost for short spans2. Standard details available3. No shoring required

1. Limited span capabilities2. Tendency to maintenance problems3. Not aesthetically pleasing



Figure 5-4. Prestressed – Precast, I-Beam Slab Span (see following explanatory notes)(Online users can click here to view this illustration in PDF.)


Economical and Practical Span Lengths (ft)Beam Type

A B C IV VI (MOD)Economical Limit 45 55 75 115 150Practical limit 60 80 90 135 175

Approximate Superstructure Depth (ft) *Beam Type

Span (ft) A B C IV VI (MOD)45 3.167 3.66755 3.667 4.16775 4.167 5.333

115 5.333125 5.333 6.833135 5.333 6.833150 6.833

* For skews up to 65°; Structure depths include 2" estimated beam haunch

Advantages Disadvantages1.Invariably the most economical for spans

between 45 ft. and 145 ft.2.Design computerized and beam details

standardized3.Adaptable to most geometric conditions

1. Not a minimum depth structure2. Cannot be curved or cut to fit extreme geometry3. Long beams sensitive to handling stresses



Figure 5-5. Prestressed – Precast, Double Tee Beam Spans (see following explanatorynotes) (Online users can click here to view this illustration in PDF.)


Approximate Superstructure Depth *Span (ft) Beam Type Depth (ft)

30-35 T21 or 22 2.1730-50 T27 or 28 2.7530-60 T35 or 36 3.42

* For skews up to 30°

Advantages Disadvantages1.Probably the most economical precast bridge in

the 30 ft. to 40 ft. span range2.Minimal diaphragm formwork to remove

underneath

1.Overlay thickness varies with camber2.Not appropriate for flared or curved structures

or skews



Figure 5-6. Prestressed – Precast, TxDOT Box Beam Spans (see following explanatorynotes) (Online users can click here to view this illustration in PDF.)


Approximate Superstructure Depth (ft) 1

Beam TypeSpan (ft) 20" 28" 34" 40"

35 2.08365 2.083 2.75080 2.750 3.250

100 3.250 3.750115 3.750

1 For skews up to 45°; however, all skews are discouraged. Depths include 5" reinforced concrete slab. Useof 2" asphalt overlay, instead of the 5" reinforced concrete slab, decreases the structure depths.

Economical and Practical Span Lengths (ft) 2

Beam Type20" 28" 34" 40"

Economical LimitPractical Limit 65 80 100 115

2 For spans utilizing 5" concrete slab and shear keys. Use of 2" asphalt overlay, instead of the 5" reinforcedconcrete slab, decreases the economical and practical span lengths.

Advantages Disadvantages1.Absolute minimum depth of precast bridge for

short and intermediate spans2.Expedites stage construction

1. Difficult to manufacture2. Not economical3. Subject to longitudinal and transverse cracking4. Not aesthetic5. Not appropriate for curved or flared structures6. Complicated for skews



Figure 5-7. Prestressed – Precast, Slab Beam Spans (see following explanatory notes)(Online users can click here to view this illustration in PDF.)


Approximate Superstructure Depth (ft) *Span (ft)

30 1.41755 1.667


Advantages Disadvantages1.Absolute minimum depth of precast bridge for

short spans2.Low-cost alternative for off-system short span

bridges3.Speed of construction

1. Not appropriate for curved or flared structures2. Complicated for skews



Figure 5-8. Prestressed, Continuous Slab Units (see following explanatory notes) (Onlineusers can click here to view this illustration in PDF.)


Approximate Superstructure Depth (ft)Interior Span (ft) ¼ Point Bent

60 1.500 3.00080 2.000 4.000

100 2.500 5.000

Advantages Disadvantages1.Absolute minimum depth for intermediate spans2.Can be aesthetically pleasing3.Deflections controlled by prestressing

1. Design and detailing more complicated2. Slightly less economical than constant depth3. Requires falsework for form support



Figure 5-9. Prestressed – Precast, U-Beam Spans (see following explanatory notes) (Onlineusers can click here to view this illustration in PDF.)



Span (ft) U40 U54100 4.050110 4.050120 5.210130 5.210


Advantages Disadvantages1. Can be aesthetically pleasing2. Easier to cast than box beams

1. More difficult to transport2. Expensive compared to prestressed I-beams



Figure 5-10. Segmental, Continuous Box Girder Units (see following explanatory notes)(Online users can click here to view this illustration in PDF.)


Approximate Superstructure Depth (ft)Max Span (ft) Skew

Approximately 0°°°°160 7.333 Structure type should be carefully180 7.667 considered during the design of200 8.000 highway alignment to optimize240 9.600 constructibility.300 12.000

Advantages Disadvantages1.Aesthetically pleasing2.Can be erected with minimum interference

beneath structure

1.Possibly uneconomical2.Awkward for flaring roadways3.May be difficult to fabricate and erect unless

methods and procedure are carefully planned4.Horizontal curvature limited to about 4°



Figure 5-11. Segmental, Simple Span Box Girder Units (see following explanatory notes)(Online users can click here to view this illustration in PDF.)


Approximate Superstructure Depth (ft)Max Span (ft) Skew

Approximately 0°°°°80 6.167 Structure type should be carefully

100 6.167 considered during the design of120 6.583 highway alignment to optimize140 6.917 constructibility.160 7.333

Advantages Disadvantages1.Aesthetically pleasing2.Can be erected with minimum interference

beneath structure

1.Possibly uneconomical2.Awkward for flaring roadways3.May be difficult to fabricate and erect unless

methods and procedure are carefully planned4.Long-term performance unproven5.Not practical for small projects6.Horizontal curvature limited to about 8°



Figure 5-12. Steel, I-Beam Units (see following explanatory notes) (Online users can clickhere to view this illustration in PDF.)



Span (ft) W18 W21 W24 W27 W30 W33 W3640 2.33350 2.58360 2.83370 3.08380 3.33390 3.583

100 3.833* For skews up to 70°

Advantages Disadvantages1.Easier connections and shaping to unusual

geometry2.Can be aesthetically pleasing

1.Expensive, except in comparison to prestressedconcrete box beams

2.Painting is unreliable, possibility of corrosion



Figure 5-13. Steel, continuous Plate Girder Units (see following explanatory notes) (Onlineusers can click here to view this illustration in PDF.)


Approximate Superstructure Depth (ft)Alignment

End Span (ft) Tangent 5°°°° Curve 10°°°° Curve100 4.000 4.000 4.250140 4.500 4.750 5.000180 5.750 6.000 6.500220 6.500 6.750 7.250260 7.500 7.750 8.250

Advantages Disadvantages1. Usually the best choice for spans over 145 ft.2. Can be curved or cut to any geometry3. Lighter than concrete superstructures4. Can be aesthetically pleasing

1.Expensive2.Painting is unreliable3.Weathering steel stains supports and rusts under

continuous moisture or salt exposure and maynot reach desired appearance in extremely dryclimate



Figure 5-14. Steel, Continuous Trapezoidal Girder Units (see following explanatory notes)(Online users can click here to view this illustration in PDF.)


Approximate Superstructure Depth (ft)Alignment

End Span (ft) Tangent 5°°°° Curve 10°°°° Curve100 This table is under review and will be140 completed at a later date.180 Contact the TxDOT Bridge Division,220 Bridge Design Section for information.260

Advantages Disadvantages1.Aesthetically pleasing, alone or with concrete

U-Beam approaches2.Erection stresses less than for I-Girders3.Corrosion possibilities less than for I-Girders

1.Very expensive2.Weathering steel stains supports and rusts under

continuous moisture or salt exposure and maynot reach desired appearance in extremely dryclimate

3.Large splices4.Heavier sections to transport and erect5.Two-girder units are considered fracture critical

Types of Superstructure and Quantity Breakdown ofOn-System Bridges in Texas (FY 2000)

Superstructure Type Number ofBridges

Concrete Slab 3,496Concrete Girder 1,172Pan Form Girder 3,688Timber 25Prestressed Concrete Beam 6,532Prestressed Concrete Box Beam 691



Steel I-Beam 2,640Steel Plate Girder 804Steel Truss 57Concrete U-Beam 29Segmental Concrete Box Girder 47Cable Stayed 2Movable 9Other Types 133Total 19,325

Bridge Class Culverts (Culverts > 20 ft.) 12,996

Substructure

Most structures with round columns and rectangular caps are satisfactory for bridges in ruraland some urban areas. However, for bridges in urban areas, especially multilevelinterchanges, aesthetics is an important factor to consider. A discussion on aesthetics can befound in Section 4 of this chapter.

Caps shaped like an inverted tee are often used to reduce the amount of cap and number ofcolumns visible beneath the superstructure where fewer, wide, rectangular-shaped columnsreplace slender circular columns. If the roadway width and span lengths are not too great, itmay be advantageous to use single column bents.

Round or square columns usually rest on single drilled shafts or a footing that caps a groupof piling. For short bents on stream crossings, a line of piling may be extended into the cap,forming a trestle pile bent. Single columns usually rest on a footing that caps a group ofdrilled shafts or piling.

Foundation

Drilled shafts are favored for bridge foundations in most areas of the state. Drilling andfounding conditions are good except for districts in the coastal plain. Using the slurrydisplacement method, it is often feasible to construct drilled shaft foundations even in veryweak soils.

♦ Spread footing foundations are seldom used and should never be used in an erodablestreambed.

♦ Concrete pilings are used extensively in the coastal areas and occasionally in otherlocations. They may compete economically in trestle pile bents even if the soil issuitable for drilled shafts. Currently, concrete pilings are always prestressed.

♦ Steel H Pilings are used sparingly. They are easier to drive than concrete pilings, andthey develop the required resistance more consistently when tipped in firm material.Corrosion possibilities make steel less desirable for trestle piling.



♦ Occasionally, some application will be found for timber piling, such as for dolphins orfender systems.

Table 5.4 contains a summary of the quantities of foundation types let to contract by TxDOTduring the fiscal year ending December 1999, which will indicate relative usage.

Foundation Types and Usage in Texas (calendar year1999)

Foundation Type Quantity (ft.)Drilled Shafts 466,300Concrete Piling 282,000Steel H Piling 11,200Timber Piling 4,500

Chapter 5 — Preliminary Considerations Section 3 — Economics


Section 3 Economics

Economics is not always the only basis for selecting structure type. Hydraulics and/orgradeline constraints may call for extra-thin superstructures. Structures that can beconstructed rapidly might be justified if detour time can be minimized. Environmentalconsiderations could justify the extra cost of especially aesthetic structures.

Texas has long been favored with low bridge costs compared to the national average. Thishas been attributed to consistency of details, good competition among the contractors, andreasonable labor costs.

Many things must be considered in planning a highway project. The planners shouldcontinue to emphasize economy.

Chapter 5 — Preliminary Considerations Section 4 — Aesthetics


Section 4 Aesthetics

It is the responsibility of the bridge designer to ensure that the structure meets therequirements for safety, durability, and cost. The engineer is equally obligated to considerthe aesthetic impact of the bridge at its proposed location. Often, a utilitarian and plainstructure efficiently satisfies the functional need. Increasingly, however, there is societalpressure to design a bridge that more harmoniously blends into and complements theparticular site with attention given to the local architecture styles and culture.

The phrase that is most often used currently is context sensitive design, which simply pointsto a site-specific design integrating both function and aesthetics. While a particular designmay take into account the local architecture close to the project, there are other designopportunities to propose a bridge that is unique to a site that otherwise has no nearbyarchitecture from which to be inspired. As an example, consider a remote and lengthy watercrossing where the future adjacent architecture may be influenced by that of the new bridge.Aesthetic decisions made at the time of the design must also take into account theanticipated adjoining land use during the life of the structure (generally taken as 75 to 100years). A rural bridge today might be the focal point of a residential or resort area in 50years.

The key to a successful aesthetic project is to build a unique and noteworthy structure —without breaking the bank. The team approach is recommended to determine the best stylefor the site, but as a bridge is an engineered structure, engineers must make the finaldecisions. The design team should take into account the input of bridge engineersexperienced and accomplished in aesthetics, architects, landscape architects, artists, andother interested parties. Other valuable input has come from the public meetings process,where users, neighbors, and landowners are asked to participate and give their opinions.Again, the owner/engineer must take the responsibility for the final aesthetic decisions.

Bridges, by virtue of their size alone, are often works of public art. In consideration of theiranticipated structural life of 75 years or more, they leave a lasting testimonial to the spiritand the priorities of those whose decisions shaped their look.

Chapter 5 — Preliminary Considerations Section 5 — Corrosion Problems


Section 5 Corrosion Problems

Information for this section will be added later. For information, contact the TxDOT BridgeDivision’s Construction Section.

Chapter 5 — Preliminary Considerations Section 6 — Railing


Section 6 Railing

There have always been many and various opinions about how a bridge railing should lookor how strong it should be. If the Bridge Design Section standards were not promoted soaggressively, there might be a different set of designs in each district.

A summary of bridge railings let to contract in Texas during the year ending December 1999is shown in Table 5.8.

Table 5.8: Bridge Railing Usage on 12 Monthly Lettings EndingDecember 1999

Railing Type Quantity(L.F.)

Number ofJobs

T501 and T501 (MOD) 388,200 66T501 Retrofit 15,900 13T502 and T502 (MOD) 131,200 51T502 Retrofit 4,000 4SSTR 169,300 10T101 and T101 (MOD) 13,800 34T201 and T201 (MOD) 10,900 8T202 and T202 (MOD) 43,000 46T6, T6 (MOD), and Retrofit 30,900 77Pedestrian and Bicycle 20,500 36Combination 48,100 39Miscellaneous 39,100 53Total 914,000 437

A new set of crash test requirements for bridge rails have been imposed by the FHWA.NCHRP Report 350 covers these requirements. Testing and updated standard developmentis currently ongoing.

The 500 series railings are 32 in. high concrete walls with the safety shape face.

A variation of the 500 series, having a straight slope on the traffic side, has been developed.It is called the Single Slope Traffic Rail (SSTR) and is 36 in. tall.

Type T4(S) is a combination concrete parapet and steel top rail with an overall height of 33in. Improved visibility and aesthetics make this rail a popular choice in urban areas. Thisrail has been successfully crash tested under NCHRP criteria.

Type T4(A) is similar to T4(S) but with an aluminum top rail.

Type T202 is a 27 in. high open concrete railing worthy of consideration. Visibility over therailing by vehicle occupants is good. The openings facilitate snow removal and provide

Chapter 5 — Preliminary Considerations Section 6 — Railing


relief from overtopping floods, as well as enhancing driver comfort. This rail is currentlybeing modified with a wider top beam. The new configuration has passed crash tests forcompliance with NCHRP Report 350. The standard designation for the new version will beT203.

Type T201 is a 27 in. high vertical wall that meets NCHRP Report 350 criteria.

Type T101 is a 27 in. high strong steel post railing system that continues the approachguardrail continuously across the structure.

Type T6 is a 27 in. high tested weak steel post system with strong tubular W-beam railmember.

The standards also contain retrofit details, pedestrian, combination traffic and pedestrian,and bicycle railings that are suitable for most conditions with occasional modifications.

Designers should choose a railing from the available TxDOT standards and consider thatthere could be a configuration more suitable than the safety shape for a given condition.Standards for rails currently used in Texas are available on the TxDOT web site.

For a more complete discussion on bridge railing, refer to the Texas Bridge Railing Manual. 1 “Splices and Anchorage of Reinforcing Bars,” Furlong, R.W. and others, CFHR, Reports 113-1, 113-2, 113-3,113-4, and 113-5F, 1968 to 1971.“Factors Affecting Splice Development Length,” Orangun, C.o., and others, CTR, Reports 154-1, 154-2, and 154-3F, 1975.“Influence of Casting Position and of Shear on the Strength of Lapped Splices,” Breen, J.E. and others, CTR,Reports 242-1, 242-2, and 242-3F, 1981.


Chapter 6 General Design Controls

Contents:Section 1 — Specifications ...................................................................................................6-2

Section 2 — Loading ............................................................................................................6-4

Section 3 — Load Distribution ...........................................................................................6-13

Section 4 — Design Methods .............................................................................................6-16

Section 5 — Design Philosophy .........................................................................................6-18

Chapter 6 — General Design Controls Section 1 — Specifications


Section 1 Specifications

AASHTO Specifications

The Standard Specifications for Highway Bridges1 adopted by the American Association ofState Highway and Transportation Officials (AASHTO) controls the design of bridges andculverts under highway traffic. These specifications were first published in 1931. They havebeen revised and republished approximately every four years since then. In the past 40years, tentative or interim revisions have been published annually that carry the full force ofthe specifications. Guide specifications have been published that have the status ofsuggested or trial specifications. After a few years of use and necessary revisions, they aregenerally incorporated into the regular specifications. Separate specifications and manualshave also been published for unusual types of structures or particular areas of bridgemanagement.

A supplement2 to the 1944 AASHO Design Specifications entitled THD Supplement No. 1was issued originally on May 24, 1945. It contained 18 items that had approval datesbetween October 14, 1944, and the issue date. The supplement was revised and reissued onJune 13, 1946, this time with 17 items. The items pertaining to live load did not appear tochange during this time frame.

THD Supplement No. 1 called for reduced axle loads in the design of concrete slabs. Manystructures designed during the era will not have an acceptable rating and are usually replacedrather than rehabilitated or widened. Careful consideration should be given to structureswhere THD Supplement No. 1 is referenced in the plan notes.

AASHTO Specifications are proposed, discussed, and approved by the AASHTO HighwaySubcommittee on Bridges and Structures, which is composed of the 50 state bridgeengineers and representatives from the Federal Highway Administration, Puerto Rico,Guam, Mariana Islands, and seven Canadian provinces. The membership is divided amongapproximately 20 technical committees, each responsible for a certain area of thespecifications. These committees continually monitor their specification areas, considersuggestions for change from other sources, and present any needed revisions to the fullsubcommittee for consideration and possible approval for AASHTO publication. Revisionsmay be suggested by users of the specifications, researchers, or organizations representingindustries that supply the various bridge materials or components. Industry proposals areusually based on the latest research in the area.

FHWA and Industry Participation

The Federal Highway Administration (FHWA) sits in review of state practice involving theuse of federal funds. They are often in the position of strongly advocating specificationrevisions, usually perceived to respond to national safety concerns. FHWA has been knownto enforce its own specifications on the states in sensitive areas but currently appears to befollowing AASHTO-approved specifications and manuals.

Chapter 6 — General Design Controls Section 1 — Specifications


Industry participation in maintenance of the AASHTO Specifications is very important.Research sponsored by organizations representing suppliers and fabricators of concreteproducts, reinforcing steel, prestressing systems, structural steel, culvert pipe, timber, andaluminum is the basis for much of the current specifications. Research sponsored byAASHTO through the National Cooperative Highway Research Program (NCHRP) has alsofurnished valuable background. The FHWA directly funds research that often finds its wayinto the specification. Research sponsored by individual states using Highway Planning andResearch funds has been useful. Texas conducts a Cooperative Highway Research Programwith state universities using these funds.

Recent Changes

Although research drives many of the specifications, there are often compromises workedout during committee deliberations. The rationale behind many provisions is lost due to lackof written commentary. Some of the early provisions, reflecting the wisdom of dominantbridge engineers of that time, still remain in the specifications. Lately, through an NCHRPproject, the AASHTO Specifications have been completely rewritten using currentknowledge and specification logic. A complete commentary is provided in the new versionas a historical record of specification changes. This commentary is found in the “Load andResistance Factor Specification,” which has the status of a guide specification in Texas.

While generally bound to compliance with the AASHTO Specifications, Texas designpractice departs from it in a few areas. Such departures are based on proven experience orlocal research. Some of these areas will be identified in the following sections of the manual.


Design of structures to carry railroad traffic is controlled by the American RailwayEngineering and Maintenance-of-Way Association (AREMA), Manual for RailwayEngineering.3 Additionally, some railroad companies have expanded interpretations orprovisions that must be followed for structures supporting their trains. These exceptions willbe addressed in Chapter 9, Section 5 of this manual.

Wind-Sensitive Structures

Wind-sensitive structures are subject to the AASHTO Specifications for Structural Supportsfor Highway Signs, Luminaires, and Traffic Signals. These structures will be discussed inChapter 9, Sections 19, 20, and 21 of this manual.

Chapter 6 — General Design Controls Section 2 — Loading


Section 2 Loading

Overview

Most short span bridges can be adequately designed using only dead and live loads. Deadload is simply the weight of the structure. Live load is whatever the governing specificationrequires for service loads to be resisted by the structure throughout its life.

The loading of bridges and structures associated with bridges, such as sign supports, isdiscussed in several mandatory specifications, as discussed in Chapter 3 of this manual. Themost commonly applicable specifications include the following:

♦ AASHTO Standard Specifications for Highway Bridges

♦ AREMA Specifications

♦ AASHTO Standard Specifications for Structural Supports for Highway Signs,Luminaires, and Traffic Signals

The loading criteria presented in each of these specifications are mandatory for theappropriate structures covered by each. This section provides some additional informationconcerning the Texas Department of Transportation (TxDOT) policy for each of thesespecifications.

Additionally, loads on bridges are thoroughly discussed by the American Society of CivilEngineers4 (ASCE). Although not directly applicable to Texas bridge design, this referencecan be used as further guidance.

AASHTO Standard Specifications for Highway Bridges

Early Texas specifications required structures to safely carry 125 pounds per square foot as alive load or a 20 ton roller, whichever required the greater strength. The 1931 AmericanAssociation of State Highway Officials (AASHO) specifications established the H10, H15,and H20 truck loadings, truck trains, and equivalent lane loadings that remain in effecttoday. The 1941 third edition added the current group of H-S truck trailer loads with anequivalent lane loading heavier than for H loads. The 1944 version made the equivalent laneload the same for H and H-S loading. A military loading for interstate highways wasintroduced by the Bureau of Public Roads in 1956.

Highway loads have increased in size and frequency during the past 50 years, but the designload has remained virtually the same. The effects of a 36 ton HS20 design load are generallya little more severe than the current 40 ton legal 18-wheeler because of the number andspacing of the rear axles. The trucking industry continually seeks to raise the legal load andsize limits. A few states design for an HS25 loading, but the American Association of StateHighway and Transportation Officials (AASHTO, formerly AASHO) has not seen fit torequire this. Texas has used HS25 for some bridges in the Texas-Mexico border area whereheavier loads due to international truck travel are encountered. HS25 is also being



considered for new bridges on North American Free Trade Agreement (NAFTA) truckcorridors. If justified, other areas in Texas may use HS25 upon approval from the BridgeDivision. Some states require their bridge designs to safely carry a family of overload truckconfigurations permitted over their highways. This practice is recognized by AASHTO.

Application of Live Loads. Texas generally uses only the live loadings prescribed by theAASHTO Specifications, applied as shown in Figure 6-1 and Figure 6-2. As discussedabove, there may be instances where alternative loadings are considered. Bridges on allhighway systems are currently being designed, at the minimum, for the HS20 loading andalso for the military loading. The military loading only controls span lengths up to 37 ft. anddoes not apply to deck slabs and direct traffic box culverts.

When applying live load, the following guidelines should be followed:

♦ Specified design live loads are placed in each traffic lane as necessary to causemaximum stress.

♦ Only one design truck per lane is placed in a span or unit.

♦ Equivalent lane loads are placed in spans as necessary to produce maximum stress. Forcontinuous spans, the lane loading shall be continuous or discontinuous as to produce amaximum value, and a second concentrated load is used to produce maximum negativemoment. Refer to AASHTO’s Standard Specifications for Highway Bridges foradditional information concerning continuous spans.

There has been research and statistical analysis directed toward a realistic mix of vehicleloads for various types of bridges. These deliberations are very complex but it is reportedthat the AASHTO lane loading may be unrealistically severe for long span bridges. Becauseof this, revised loadings are occasionally negotiated for long span bridges.

Impact Due to Live Loads. Live loads shall be multiplied by an impact factor to increasethe effects of the live load to account for effects due to vibration and impact per AASHTOSpecifications. For the analysis of the structure, the effects of impact shall be transferredfrom superstructure to substructure but shall not be included in loads transferred to structuralelements below the ground line for the analysis of those structural elements.

Application of Other Loads. In addition to dead loads and live loads, the followingAASHTO Specifications loads are common to bridges:

♦ Centrifugal Force. Centrifugal forces due to live load may be treated as shown inFigure 6-3.

♦ Longitudinal Force. Longitudinal forces due to live load are thoroughly described inAASHTO Specifications.

♦ Wind Load. Wind loads must be considered but will seldom control the design of gradeseparation or stream crossing structures less than 25 ft. above the ground.

The 1931 specification required bridges to resist a wind load of 30 pounds per square footon 1½ times the area as seen in elevations, plus all girders in excess of two in the crosssection. The origin of this loading is lost in antiquity. The load was changed to 50 poundsper square foot on 1½ times the area in 1953 and then to the current 50 pounds per square



foot on the area as seen in elevation in 1957. Trusses and arches are designed for 75 poundsper square foot.

The AASHTO Standard Specifications for Structural Supports for Highway Signs,Luminaires, and Traffic Signals5 contains a more refined treatment of wind loads. There isan excellent treatise on wind with a large bibliography in the ASCE Transactions, Paper No.3269.6

If the structure is considered sensitive to wind, the forces are applied according to Figure 6-4. Wind on the live load is also covered on this figure.

Long span structures may justify more sophisticated analyses, including wind tunnel tests, toinvestigate the dynamic performance of the design.

Other loads mentioned in the AASHTO Specifications are treated as follows.

♦ Electric Railway Loads. Electric railway loads are a holdover from early specifications,but streetcars are becoming popular again.

♦ Sidewalk Loading. Sidewalk loading shall be applied as described in the specifications.

♦ Curb Loading. Curb loading shall be applied as described in the specifications. Curbsare seldom used on bridges.

♦ Railing Loading. Current railing standards are designed to AASHTO requirements, butthe trend is toward crash testing to verify railing details.

♦ Stream Current. Stream current should be considered but rarely controls the design.Designing for drift loads is highly speculative. If significant drift is expected, wall orwebbed piers should be used and careful attention given to span lengths and skew angle.

♦ Ice Pressure. Ice pressure does not occur in Texas.

♦ Buoyancy. Buoyancy is important for cofferdams but is seldom a factor in ordinarybridge design.

♦ Earth Pressure. Use an equivalent hydrostatic pressure of 40 pounds per cubic footunless more exact determinations are justified.

♦ Earthquake Motions. At the present time, TxDOT does not design for earthquakes.

♦ Temperature, Shrinkage, and Rib Shortening. These factors are listed in thecombination of loads, but they are actually internal deformations that can result in stressredistribution. Temperature deformations, as defined in the specifications, areconsidered in the design of substructure for continuous units. Shrinkage is seldomtreated analytically. Rib shortening is a secondary effect that should be considered inthe design of arches.Temperature, shrinkage, and creep have been observed to have a significant effect onlarge concrete box girders. Research has verified some of the parameters and methodsavailable to analyze their effects. All concrete box girders should undergo such ananalysis.

Load combinations. Seven combinations of the above loads (Groups I through VI and X inthe AASHTO Specificaitons) must be considered in the design of bridges.



Figure 6-1. AASHTO HS20 Truck, H20 Truck, and Alternate Military Live Loads (seefollowing explanatory notes) (Online users can click here to view this illustration in PDF.)


Figure 6.1 shows equivalent lane loading for HS20 and H20 trucks. When applying the liveload to the design, remember the following:

♦ Reduce live load effects by 10 percent if three lanes are loaded.



♦ Reduce live load effects by 25 percent if four or more lanes are loaded.

♦ Increase for impact per AASHTO Specifications.

Regarding 18k for moment and 26k for shear notes: An additional concentrated load is usedin the design of negative moment regions for continuous spans.

Regarding 0.640 klf notes: In the design of continuous bridges, the uniform load is placed inspans only as necessary to produce maximum stress.

Alternate military loading, developed by the FHWA in 1956, represents heavy militaryvehicles.

All bridges on the U.S. Interstate Highway System, or any highway bridge that may carryheavy truck traffic, are to be designed using HS20 or the alternate military loading,whichever produces the greatest stresses.



Figure 6-2. Applying Live Load on the Structure for Slab, Beam, and Bent Design (seefollowing explanatory notes) (Online users can click here to view this illustration in PDF.)

Explanatory Notes for Figure 6.2

LL = 2P20 + 10w; Live load reaction per land; controlling between truck and land load,increase for impact if applicable.

P20 = 16k; The load on one rear wheel of HS20 or H20 truck, increase 30 percent for impactif applicable (1.3P20 = 20.8k).

102PLLw 20−=

The uniform load portion of LL (k/ft).

Slab Design. Specific slab design moments and distribution widths are specific byAASHTO. Uniform load (w) is not applicable and is ignored. Wheel load (P20) shall beincreased 30 percent for impact (1.3P20 = 20.8k). Exterior P20 shall be placed 1 ft. 0 in. fromface of rail when designing cantilever.

Beam Design. Use live load distribution factors specified by AASHTO.

Bent Design. The live load is distributed to the stringers assuming the slab is simplysupported at each beam. The live load is placed at critical locations, and using combinationsof loaded lanes as to produce the maximum stresses. Wheel load (P20) shall be increased 30percent for impact (1.3P20 = 20.8k) for substructure elements above the ground line.



Figure 6-3. Application of Centrifugal Force (CF) (see following explanatory notes)(Online users can click here to view this illustration in PDF.)


Centrifugal Force (CF) = RF (n)(C)LLTL)Where:

RF = Reduction in load intensity factor; applicable only if n ≥ 3, per AASHTON = Number of loaded lanesC = Centrifugal force in percent of live loadLLTL = Live load due to truck load without impact, kips (k)

The direction of CF is radial. If the bent is skewed, the radial force shall be resolved intoparallel and perpendicular components.

The centrifugal force in percent of live load shall be calculated by the following:

C = 0.0000117S2DWhere:

S = Design speed in mph, if no superelevation is presentD = Degree of curvature along the baselineTo account for the effects of superelevation, truck speed (S) may be taken as the following:

( )0.15eD

85950S += where e = superelevation rate



Figure 6-4. Application of Wind Loads, including WSUP, WSUB, WUP, WL (see followingexplanatory notes) (Online users can click here to view this illustration in PDF.)


♦ Wind on superstructure (WSUP) is 50 psf transverse, applied simultaneously with 12 psflongitudinal and resolved into components parallel and perpendicular to the bent.

♦ Wind on the substructure (WSUB) is 40 psf transverse, applied simultaneously with40 psf longitudinal and resolved into components parallel and perpendicular to the bent.

♦ Uplift (WUP) is 20 psf of deck and sidewalk plan area applied at the windward ¼ pointof the transverse superstructure width.

♦ Wind on live load (WL) is 100 plf transverse, applied simultaneously with 40 plflongitudinal and resolved into components parallel and perpendicular to the bent.




The American Railway Engineering and Maintenance-of-Way Association specifications forloading are strict, but can be followed without any undue expense. Refer to the AREMAspecifications for loads and methods of application.

AASHTO Standard Specifications of Structural Supports for Highway Signs, Luminaires,and Traffic Signals

Used for the design of sign supports and poles, these specifications frame a different type ofstructural design for which wind speed, ice load, and shape factor are importantconsiderations. Refer to the AASHTO Standard Specifications for Structural Supports forHighway Signs, Luminaires, and Traffic Signals for loads and methods of application.

Chapter 6 — General Design Controls Section 3 — Load Distribution


Section 3 Load Distribution

Overview

Truck wheel loads are delivered to a flexible support through compressible tires, whichmakes it very difficult to define the area of the bridge deck significantly influenced.Computerized grid systems and finite element programs can come close to reality, but theyare complicated to apply and are limited by mesh or element size and by the accuracy withwhich the mechanical properties of the composite materials can be modeled. These two- orthree-dimensional problems are reduced to one dimension through various empiricaldistribution factors given in the AASHTO Specifications.

These distribution factors have been derived from research involving physical testing and/orcomputerized parameter studies. In order to simplify the design procedure, the number ofvariables was reduced to a minimum consistent with safety and reasonable economy,according to the judgment of the AASHTO Highway Subcommittee on Bridges andStructures. The factor S/5.5, so developed, has been used for many years to determine theportion of a wheel load to be supported by steel or prestressed concrete stringers under aconcrete slab. Other variables, such as span aspect ratio, skew angle and relative stiffnessbetween stringer and slab, are not considered except for occasional special bridges. Theconservatism of this approach may account for some of the reserve strength regularlyobserved when redundant stringer bridges are load tested. Similarly, experience has shownthat concrete slab spans and slabs on stringers will invariable support much more load thanpredicted by empirical analysis.

Load Distribution

Treatment of wheel load distribution to the various bridge components in the AASHTOSpecifications is as follows:

♦ Longitudinal Beams (Stringers). Distribution factors given in the specifications areused almost exclusively. Occasionally, special conditions will justify the use of adiscrete element grid and plate solution.

For simplicity of calculation and because there is no significant difference, thedistribution factor for moment is used also for shear. Composite dead loads aredistributed equally to all stringers except for extraordinary conditions of deck width orratio of overhang to beam spacing. Live load is distributed to all types of outside beamsassuming the deck to act as a simple cantilever span supported by the outside and thefirst inside stringer.

♦ Transverse Beams (Floorbeams). For the few cases where floorbeams have been usedwithout stringers on highway bridges, it has appeared proper to calculate reactionsassuming the deck slab to act as a continuous beam supported by the floorbeams. Notransverse distribution of wheel loads is allowed unless a sophisticated analysis is used.



♦ Concrete Slabs - Reinforced Perpendicular to Traffic (Slab on Stringers). For thiscomponent, distribution of the wheel load is built into a formula for moment. TxDOTdesigns are standardized according to the requirements of the current AASHTOSpecifications. Span length of slabs on prestressed concrete stringers may be taken asthe clear distance between flanges and adjusted to the flange quarter points for steelstringers.

♦ Slab Overhang Design. This design is also standardized.

♦ Concrete Slabs - Reinforced Parallel to Traffic (Slab Spans). Loads are distributedaccording to the AASHTO Specifications. The approximate formula for moment is notused.

For skews up to 30 degrees, main reinforcing is parallel to traffic, and no additionaledge beam strength is needed for usual railing conditions. For skews greater than 30degrees, reinforcing is perpendicular to the bents and edge beam strength is providedand reinforced parallel to traffic.

♦ Concrete Slabs - Reinforced Both Ways. Divide the load between transverse andlongitudinal spans according to the formulas for slabs supported on four sides. Use theappropriate load distribution in each direction.

♦ Timber Flooring, Composite Wood-Concrete Members, and Glued Laminated TimberDecks. Timber is not used in new structures.

♦ Steel Grid Floors. The specifications are followed closely. This type of construction isseldom used in Texas.

♦ Spread Box Girders. The specifications are followed closely. This type of constructionis seldom used in Texas.

♦ Precast Concrete Beams Used in Multibeam Decks (Box Beams). The latest standarddesigns and current special designs comply with the current AASHTO Specifications.The distribution factor is a function of box width, overall bridge width, number of lanes,and span length. The simplified values for K shown in the specifications are usuallyused for final designs.

♦ Other Structure Types. See Chapter 7 for distribution factors for other structure typesnot listed here.

Horizontal Loads

Horizontal loads on the superstructure distribute to the substructure according to acomplicated interaction of bearing and bent stiffness. For continuous steel units, thefollowing method will usually be sufficiently accurate:

♦ Apply transverse loads times the average adjacent span length.

♦ Apply longitudinal loads times the unit length to the fixed bents according to theirrelative stiffness.

♦ Calculate deformations due to temperature change of 70 degrees and convert to forcesaccording to the stiffness of the fixed bents.



♦ Centrifugal force is based on the truck load reaction to each bent.

♦ Friction in expansion bearings can usually be ignored but, if its consideration isdesirable, the maximum longitudinal force may be taken as 0.10 times the dead loadreaction for rocker shoes and polytetra fluoroethylene (PTFE) sliding bearings.

For prestressed concrete beam spans and units on elastomeric bearings, fixity is superficialand all bearings are approximately the same stiffness. It will usually be sufficiently accurateto distribute horizontal loads in the following manner:

♦ Apply transverse and longitudinal loads times the average adjacent span length. Theconcentrated live load for longitudinal force would be located at each bent.

♦ Centrifugal force is based on the truck load reaction to each bent.

♦ Forces due to temperature deformations may be ignored except for bearing design. Iftemperature consideration is desirable, deformations may be based on 40 degreetemperature change.

♦ Bearing stiffness may be based on a shear modulus of 175 psi. For a completediscussion on bearing pad design, refer to the paragraph titled “Elastomeric Bearings”in the Chapter 9 discussion of Design Recommendations for Bearings.

Chapter 6 — General Design Controls Section 4 — Design Methods


Section 4 Design Methods

Overview

Under the current AASHTO Specifications, two basic design methods are allowed —Service Load Design and Strength Design.

AASHTO Specifications

The AASHTO Specifications allow Service Load Design and Strength Design alternativelyfor reinforced concrete, structural steel, soil-corrugated metal structures, and soil-reinforcedconcrete structures. Service Load only is indicated for timber, supports for signs, luminaires,traffic signals, and railing. Prestressed concrete design uses a combination of the twomethods. Concrete strength and amount of prestressing is usually determined by ServiceLoad Analysis, but Strength Design is used for shear and the ultimate flexure check.

The Texas Bridge Design Section generally recommends Service Load Design forreinforced concrete slabs, reinforced concrete footings, and structural steel. Strength Designis recommended for reinforced concrete bent caps and columns.

Service Load. In Service Load Design, loads of the magnitude anticipated during the life ofthe structure are distributed empirically and each member is analyzed assuming completelyelastic performance. Calculated stresses are compared to specified allowables that have beenscaled down from the tested strength of the materials by a factor judged to provide a suitablemargin of safety.

Load Factor (Strength Design). In Strength Design, the same service loads are distributedempirically and the external forces on each member determined by elastic analysis. Thosemember forces are increased by factors judged to provide a suitable margin of safety againstoverloading. These factored forces are compared to the ultimate strength of the memberscaled down by a factor reflecting the possibility and consequences of constructiondeficiencies. Serviceability aspects, such as deflection, fatigue, and crack control, must bedetermined by Service Load Analysis.

The Strength Design Method produces a more uniform factor of safety against overloadbetween structures of different type and span length. Strength Design also tends to producemore flexible structures.

Load and Resistance Factor Design. A third method is proposed in the Load and ResistanceFactor Design Specification. This method has more numerous and more accurate load andresistance factors based on probabilistic theory and reliability indices and should produceeven more uniform and realistic safety factors between different types of bridges. It is, ofcourse, more complicated, which accounts for its slow acceptance by some states.

Chapter 6 — General Design Controls Section 4 — Design Methods



Design methods under the AREMA Specifications are the same as for AASHTO.

AASHTO Standard Specifications for Structural Supports for Highway Signs, Luminaires,and Traffic Signals

Design methods are different under these specifications. Determination of wind forces ismore complicated, depending on geographical location and height of the structure above theground. Moments, shears, and torsions are determined by elastic analysis, although most ofthe structures are determinate. Shape factors effect calculated stress. Although allowablestresses appear to be service load values, they are derived from load factor considerations.Wind induced oscillations, which produce fatigue stresses, must be considered in somestructures.

Chapter 6 — General Design Controls Section 5 — Design Philosophy


Section 5 Design Philosophy

Design Evolution

Long span bridges excite the imagination and bring notoriety to the owner. Texas has veryfew of these bridges. In the short and medium span categories the State owns approximately20,000 bridges not counting box culverts (see the "Superstructure and Quantity Breakdownof On-System Bridges in Texas" table in Chapter 5). At least 90 percent of these bridgeswere designed in-house. To keep up with the demand for bridge plans, it was necessary tomaintain an extensive set of standard detail sheets that could be reproduced and used in theproject plans. Many of these standard groups covered a multiplicity of roadway widths,spans and skews and could require many original drawings. An early resistance wasdeveloped to changes in roadway width and design specifications that would causewholesale standard revisions.

New structure types were developed to fill specific needs. Pan-formed concrete girders weredeveloped in the late 1940s because so many short span stream crossings were beingconstructed uneconomically with steel beams or shored concrete girders. These standardshave undergone several changes in roadway width but are still used very economically inconsiderable numbers today. Precast pretensioned beams were developed in the 1950s formedium span stream crossings and grade separations because steel beams were becomingexpensive and sometimes slow on delivery. Fewer plans are assembled from standardprestressed drawings today because bridge geometry has become so complicated andvariable that most details must be specially prepared. The beams themselves still contain thestandard shapes developed in the beginning, and the accessories required to complete thespan are covered on standard details.

New shapes have been added in concrete and steel for which standard span and bent detailshave not been fully developed. After 50 years of use, box culverts have been completelyredesigned by the load factor method and new standard drawings prepared.

Design Considerations

The Bridge Design Section has performed all types of design in-house except for cable-stayed and suspension bridges. The more advanced structure types have as yet required onlya small portion of the overall effort. The most important part of the job is design and planpreparation for multitudes of conventional bridges that usually have some variation ingeometry that prohibits the use of straight standard details.

Geometry is considered an important part of bridge design. Framing dimensions andelevations must be accurate to avoid expensive field correction. Design engineers areprimarily responsible for geometric accuracy.

Constructibility is important. There have been designs that looked good on paper but werevirtually impossible to properly construct. Designers need to consider how to build thecomponents. Construction experience is a valuable asset.



Details are more critical in design. Failure to provide for proper stress flow at discontinuitieshas often caused local distress and sometimes mortal injury to a system. Engineers andtechnicians should recognize and carefully evaluate untested details.

Maintenance Considerations

The bottom line on bridge design is maintenance. It is usually much more expensive torepair a bridge than it was to build it. Unfortunately, maintenance problems tend to occurseveral years after the structure is built. During that time there may be many more bridgesdesigned with the same problem. Experience is a good teacher, but the lesson is sometimesslow to be learned. It takes a good designer to anticipate maintenance problems and spendjust enough of the taxpayers’ money to prevent or delay problems.

Design engineers are expected to develop engineering judgment to recognize the degree ofdesign complexity and accuracy justified by the type and size of structural element underconsideration. A number of computer programs are available. Some are so complicated as tobe useful in very special investigations only. Others, although complicated, offer the onlyrealistic solution to a problem. Others are very useful and save time during design andproduction. Virtually every bridge designer has access to a personal computer that can runmost of the programs immediately. However, longhand methods may still be desirable insome areas.

Design Notes

Design notes are the documentation for structural adequacy and accuracy of pay quantitiesfor each bridge. These notes are kept on file for a reasonable period after construction of thebridge. The condition of the design notes reflects the attitude of the designer and checker.

Quantity calculations should be emphasized. For “Plan Quantity” items, the constructionspecifications state that “adequate calculations have been made in accordance withArticle 9.1.” The design office is the Office of Record responsible for the accuracy of payquantities. It is imperative that accurate calculations, which agree with the quantities shownon the plans, be found in the design note file. Plan quantity items are concrete, reinforcedconcrete slab, retaining wall, prestressed concrete beams, box beams, tee beams and slabbeams, prestressing, preformed joint seal, sealed expansion joints, concrete overlay,structural steel, and railing. Reinforcing steel, whether pay item or not, should be billed onthe plans but need not be included in the design notes. Foundation quantities should besummarized from lengths shown on the plans and not repeated in the design notes.

A Texas design memorandum issued over 25 years ago is quoted below. It is as appropriatenow as it was then.

“Our design notes should consist of a concise, but complete, clear, and easilyfollowed record of all the essential features of the final design of each structure. Itis often necessary to refer to these notes because of changes or questions whicharise during the construction period. If properly prepared and assembled, thesecalculations are of great value as a guide and time saver in preparing designs ofother similar structures.



To properly serve the above purposes, the design notes should be prepared andarranged so clearly that any reasonably experienced bridge designer can followthem and can obtain all the essential information he may need without consultingthe [person] who made them and without wasting a lot of time...

The following essential features are to be observed in preparing, checking andfiling design calculations:

The headings at the top of each sheet are to be completely filled in and each sheet isto be numbered.

The first sheet of calculations should always list clearly such governing features asroadway widths, curb or sidewalk widths and heights, and design loading. If anydeviations are to be made from the standard design specifications, these should belisted.

The first sheet of calculations on any superstructure unit should show by sketch, alayout of that unit, giving number of spans and length (c-c bearing) of each span. Aline diagram will suffice.

The first sheet of calculations of any substructure unit should show an appropriatesketch or diagram of the unit, properly dimensioned, and the superstructure shouldbe shown.

Appropriate headings and subheadings such as ‘Live Load Moments, CenterGirder,’ ‘Summary of Shears, Outside Girders,’ etc. should be freely used. Theseheadings should be supplemented by explanatory notes wherever necessary toclarify the portion of structure under consideration, the load combinations beingused, or the method of analysis being employed.

In checking calculations, do not make up a separate set of design calculations.Follow the original calculations and check them through, or at least check theresults. In those few instances where original calculations are so poorly made that anew set must be prepared, eliminate the original set and include the second set as aportion of the final calculations.

In checking calculations, don’t carry through corrections that are so minor inamount as to have no real effect on the structure. Remember that structural designis not an exact science. The live load is assumed, the wind load is assumed anddead load generally is approximate only. The strengths of the materials employedvary widely. DON’T SPLIT HAIRS. Look for 20 percent errors, not 1 percenterror.

Calculation sheets shall be placed in file in usual sequence; i.e., with Sheet No. 1on top followed in order by Sheet Nos. 2, 3, etc., instead of reverse order.Superstructure calculations shall be placed in front of substructure calculations.Quantity calculations shall be placed at the bottom of the file. Preliminary designs,trial designs and comparative designs are not to be included in the design folder asfinally filed.

Supplement the above rules with good judgment and plenty of common sense. Theextra ten minutes you spend in making your calculation sheet clear and completemay save the checker an hour, and may, two years hence, save some bridgedesigner a week or more of computations.”



1 “Standard Specifications for Highway Bridges,” American Association of State Highway and TransportationOfficials (AASHTO), Sixteenth Edition (1996).2 "Supplement No. 1 to 1944 AASHO Design Specifications for Texas Bridges (THD No. 1)", Public RoadsAdministration and Texas Highway Department, 1945.3 “Manual for Railway Engineering (Fixed Properties),” AREA 1988-89.4 “Recommended Design Loads for Bridges,” Proceedings of the American Society of Civil Engineers, Volume 107,ST 7, 1981.5 “Standard Specifications for Structural Supports for Highway Signs, Luminaires and Traffic Signals,” AASHTO,1985.6 “Wind Forces on Structures,” J.M. Briggs, Chmn., Transactions, American Society of Civil Engineers, Paper No.3269, Volume 126, Part 2, Final Report, 1961.

Chapter 7 — Superstructure Design Section 1 — One-Way Deck Slabs on Stringers


Chapter 7 Superstructure Design

Contents:Section 1 — One-Way Deck Slabs on Stringers...................................................................7-2

Section 2 — Two-Way Deck Slabs on Stringers ................................................................7-14

Section 3 — Simple Concrete Slab Spans ..........................................................................7-16

Section 4 — Continuous Concrete Slab Spans ...................................................................7-19

Section 5 — Simple Concrete Girder Spans.......................................................................7-22

Section 6 — Concrete Pan Form Slab and Girder Spans....................................................7-27

Section 7 — Continuous Concrete Girder Spans................................................................7-34

Section 8 — Concrete Box Girder Spans............................................................................7-36

Section 9 — Prestressed Concrete Deck Slabs ...................................................................7-38

Section 10 — Prestressed Continuous Slab Spans..............................................................7-40

Section 11 — Prestressed Simple Slab and Girder Spans ..................................................7-43

Section 12 — Prestressed Pan Form Slab and Girder Spans ..............................................7-44

Section 13 — Prestressed Continuous Concrete Girder Spans...........................................7-45

Section 14 — Prestressed Cast-in-Place Box Girder Spans ...............................................7-47

Section 15 — Prestressed Segmental Box Girder Spans ....................................................7-49

Section 16 — Prestressed TxDOT Box Beam Spans .........................................................7-53

Section 17 — Prestressed AASHTO/PCI Box Beam Spans...............................................7-61

Section 18 — Prestressed Slab Beam Spans.......................................................................7-63

Section 19 — Prestressed Single Tee Beam Spans.............................................................7-65

Section 20 — Prestressed Double Tee Beam Spans ...........................................................7-66

Section 21 — Prestressed Simple I-Beam Spans................................................................7-72

Section 22 — Prestressed Cantilever/Drop-In I-Beam Spans ............................................7-88

Section 23 — Prestressed Continuous for Live Load I-Beam Spans .................................7-95

Section 24 — Prestressed U-Beam Spans.........................................................................7-101

Section 25 — Rolled Steel I-Beam Spans ........................................................................7-108

Section 26 — Steel Plate Girder Spans.............................................................................7-112

Section 27 — Trapezoidal Box Girders ............................................................................7-129



Section 1 One-Way Deck Slabs on Stringers

Background

From the beginning of Texas Highway Department history, there have been reinforcedconcrete bridge decks. They were used for the entire superstructure as slab spans andculverts, placed monolithically with concrete girder spans, and constructed on top of timber,steel, and precast concrete beams.

Design Specifications

Design specifications have undergone an evolution from none in the beginning throughseveral empirical formulas for distribution of wheel loads to the completely empirical designmethod specified in the Ontario Highway Bridge Design Code. Beginning in 1980 the TexasDepartment of Transportation (TxDOT) sponsored an extensive research projectinvestigating bridge decks designed according to the Ontario empirical method. A fewexperimental bridges using the method were constructed. The Federal HighwayAdministration (FHWA) accepted this method for phasing into general use by 1991.However, after a few tentative trials, construction of this type of deck was abandoned. Thepotential savings in reinforcing steel were not significant when concrete panels were used,and the details were just more comfortable with the standard method.

Beginning with Supplement No. 1 to 1944 American Association of State Highway Officials(AASHO) Design Specifications for Texas Bridges,1 issued in 1945, the design load forconcrete deck slabs was one 24,000-pound axle or two 16,000-pound axles, 4 ft. apart. Thiswas changed after the 1957 AASHO Specification to one 32,000-pound axle.

Design Stress

Allowable stresses in the concrete have changed from 800 psi for live load in 1931 to 0.4 f 'ctoday. AASHTO-required concrete strengths have increased from 3,000 psi to 4,500 psi andback to 4,000 psi in the 1988 American Association of State Highway and TransportationOfficials (AASHTO) Interim Specification. The required 28-day concrete strength forconcrete used in bridge decks in Texas was 3,000 psi until 1974, when Class S concrete wasintroduced, which has six sacks of cement per cubic yard, low water/cement ratio, entrainedair, and a required strength of 3,600 psi. Later, in response to the 1988 AASHTOprovisions, Class S concrete was required to have six and one-half sacks of cement and astrength of 4,000 psi. Although the allowable design stress in the concrete rose to 1,800 psi,most states limited the calculated service load stress, using the empirical wheel loaddistribution, to 1,200 psi. This limit kept deck slabs from getting thinner as the concretestrength increased.



Concrete Cover

Required concrete cover to the top mat of reinforcing steel remained at 1 in. until therequirement increased to 1 1/2 in. in the 1961 Specification. The 1977 Specification required2 in., which Texas had already adopted for nominal cover. Most construction specifications,including AASHTO, allow a 1/4 in. tolerance on position of reinforcing steel, but someauthorities believe that, for top slab reinforcement, 2 in. should be the absolute minimumcover.

A chronology of dominant AASHTO effects on the design of deck slabs reinforcedperpendicular to traffic is given in the table below. Texas practice is outlined in the secondfollowing table.

Chronology of AASHTO Specification Requirements for Concrete Slabs ReinforcedPerpendicular to Traffic

Allowable Stress Concrete CoverAASHTO

Spec.DesignWheel

1 DistributionFormula

E =

Conc.(psi)

Reinf.(psi)

Top(in)

Bott.(in)

1931 16 k 0.7(S+1.25) 800 16,000 1 11935 16 k 0.6S+2.5 900 16,000 1 11941 16 k 0.4S+3.75 1,000 18,000 1 11944 12 k 0.4S+3.75 1,000 20,000 1 11953 12 k 0.4S+3.75 1,200 20,000 1 1

1959 (T) 16 k 8S/(S+2) 1,200 20,000 1 11961 16 k 8S/(S+2) 1,200 20,000 1 ½ 11973 16 k 8S/(S+2) 1,200 2 24,000 1 ½ 11977 16 k 8S/(S+2) 3 1,800 24,000 2 11983 16 k 8S/(S+2) 1,800 24,000 2 11993

toCurrent

16 k 8S/(S+2) 1,600 24,000 2 1

1. Expressed in terms of span length, for spans over 7 ft.2. Probably not used by most states.3. Construction section required 4,500 psi concrete. Probably not used in design by any state.



Chronology of TxDOT Design Requirements forConcrete Deck Slabs on Stringers

2 AllowableStress

ConcreteCover

3, 4 Slab Designfor 8 ft. Span

1 Year DesignWheel

1 DistributionFormula

E =

Conc.(psi)

Reinf.(psi)

Top(in)

Bott.(in)

E (ft) SlabThick.

(in)

TopReinf.#5 at

1946 12 k 0.4S+3.75 1,000 18,000 1.1875 1.1875 6.95 6.50 6 in.1955 12 k 0.4S+3.75 1,200 20,000 1 .1875 1.1875 6.95 6.50 6.50 in.1960 16 k 8S/(S+2) 1,200 20,000 1.1875 1.1875 6.40 6.75 5.25 in.1961 16 k 8S/(S+2) 1,200 20,000 1.8125 1.1875 6.40 7.25 5 in.1968 16 k 8S/(S+2) 1,200 20,000 2 1.25 6.40 7.75 5 in.1996 16 k 8S/(S+2) 1,600 24,000 2 1.25 6.40 8 6 in.

1 . Design practice from 1919 to 1946 cannot be documented. Beginning in 1961, no direct formula fordistribution width was given. It was simply included in the direct formula for design moment per foot.

322)(SP15 +

or 322)(SP20 +

, with a 0.8 factor allowed for slabs continuous across 3 or more supports.

2 . From 1996 to current, higher reinforcing steel stresses were allowed to justify #5 at 6 in. in the top mat.The allowable stress requirement used is the same as for prestressed concrete panel decks.

3 . Based on design clear spacing of 7 ft. (12 in. of contributing flange).

4 . For constructibility reasons, TxDOT currently recommends the use of 8 in. slabs only. However, 7.50 in.and 7.75 in. slabs may sometimes be used if it can be shown that the span can be designed using fewer beamsas the result.

Design Methods

Basic design methods were last revised in the 1959 Interim Specification. The deck slab isdesigned as a beam in flexure supported by the stringers. This design results in primaryreinforcing oriented transversely to the length of the bridge. This reinforcing is placed in twolayers within the slab. To maximize design strength, one layer is placed closest to the top ofthe slab, and the other layer is placed closest to the bottom of the slab. If the slab behavedlike a beam, cracking would be expected perpendicular to the main reinforcing orlongitudinal to the bridge. This has not been observed to occur. Cracking usually initiatedtransversely, often directly above the main reinforcing where a plane of weakness existed.Furthermore, field and laboratory tests consistently identified failure in punching shear, notflexure, at a level of six or more times the design wheel load.

Prestressed Concrete Panels

Beginning in 1963, Texas developed a deck construction method whereby about half of theslab could be precast. Prestressed concrete panels span between stringers, supporting theweight of the cast-in-place top half of the slab. Early attempts to cantilever the panels acrossthe outside beam to support the overhang encountered too many construction problems, and



the concept was abandoned. The use of prestressed concrete panels between beams onprestressed concrete beam bridges gained slowly in popularity.

When the bid item for reinforced concrete slab began to be measured by the square foot withthe specification allowing either removable forms, stay-in-place metal deck forms, or precastprestressed concrete panels, at the contractor’s option, the use of panels escalated.

Texas invested a considerable amount of time and money in research2 relative to prestressedconcrete panels. Details have been modified continually for compatibility with constructionconditions and manufacturing practice. The method is subject to grading problemsassociated with geometric conditions and prestressed beam camber. Excess cast-in-placeconcrete is usually required because of inability to grade the panels properly. A majority ofthe contractors prefer panel decks in spite of the shortcomings because a convenient and safeworking surface is provided very quickly.

Maintenance Issues

The maintenance problem with bridge decks is deterioration under exposure to weather andtraffic (see Chapter 5, Section 5:Corrosion Protection). Texas became aware of the severityof this problem after a period of extensive construction of the thinnest decks ever allowed byAASHTO. Speculation was rampant as to the causes of cracking, scaling, delamination, andfinally portions of deck falling to the ground under and over traffic. Some of the causessuggested were:

♦ Decks too thin

♦ Insufficient concrete cover over reinforcing

♦ Insufficient temperature reinforcing

♦ Slabs composite with steel beams

♦ Slabs not composite with steel beams

♦ Steel beams too limber

♦ Dirty concrete aggregates

♦ Reactive concrete aggregates

♦ Insufficient cement in concrete

♦ Water-cement ratio too high

♦ Chemical admixtures

♦ Ready-mix concrete

♦ Excessive concrete placement temperature

♦ Incomplete consolidation of concrete

♦ Slow finishing methods

♦ Lack of uniform curing



♦ Insufficient long-term curing

♦ No protective coating provided

♦ Concrete too young when traffic allowed

♦ De-icing salts

Most of these potential causes probably had some influence on the problem, and most ofthem were addressed by the various corrective measures taken in the ensuing years.

Panel decks have been used in virtually every area of the state, including where de-icingsalts are applied to the bridge deck. Other states had experienced deck deterioration causedby the use of precast panels. Initially, Texas panel decks experienced no longitudinalcracking or delamination. Transverse cracking over joints between panels occurred. Thiscracking led to changing the design standard’s longitudinal top reinforcing from #4 at 12 in.to #4 at 9 in. This change was later adopted by an AASHTO code provision. Cracks canstill be found, but their size is considered acceptable. Recently there have been reports oflongitudinal cracking in some Texas panel decks. This justifies emphasis of the requiredmethod of construction to prevent this cracking as shown in Figure 7.1.

Figure 7-1. Prestressed Concrete Panel on Bedding Strips (Online users can click here toview this illustration in PDF.)

In the middle 1970s, the design engineer and the construction engineer from Florida came toTexas to find out why their panel decks were cracking longitudinally and ours were not.They had experienced several significant failures. It was their practice to put the beddingstrips at the edge of the panel. No grout could enter to provide a firm bearing for live load.We looked at eight structures, from Cameron to Grayson counties, without finding a singlelongitudinal crack.

Deck Continuity for Simple Spans

Background. Determined to alleviate the deck joint problems with prestressed beam spans,but disenchanted with full continuity for live load, the Bridge Design Section beganexperimenting with continuity of the deck slab over interior bents. Prestressed beams wereconstructed with a 4 in. space between beam ends across the center of cap. Design of thebeams was the same as for simple spans.



On the earlier bridges, slab continuity was achieved with a single layer of dowel barsbetween adjacent spans. Success on the first bridges led to an escalation of such designs,some of which connected too many spans without proper provision for movement at theends of the continuous section. A number of structures still bear the scars of this era.

The continuous slab joint performed fairly well through all this, and gradually the dowelsgave way to continuous top and bottom reinforcing mats through the joint. Maximumcontinuous unit length criteria was established, based on experience, and expansionprovisions were insured.

Design Issues. Design in this case was by trial and error. The continuous slab across thejoint between adjacent beams cannot be justified by rigorous design methods.

Construction Issues. Construction was simplified. No diaphragms were required under thecontinuous slab at the beam ends. Slab reinforcing was less complicated on skewed bridgesbecause variable length skewed end reinforcing was not required at the continuous spanends. Expansion joint hardware was minimized.

Originally, at the contractor’s option, the slab concrete could be placed continuously acrossthe centerline of bent. Optional construction joints were given at centerline of bent and atmidspan for spans in excess of 100 ft. In the mid 1980s this practice was changed aftersurveying several bridge decks constructed with continuous placement. Irregular crackingpatterns were observed, particularly at skewed bents. In some areas a single crack wouldsplit apart and come together again leaving portions of the deck encircled by the crack.Durability concerns for these regions of the deck led the Bridge Design Section to requireforced construction joints at the centerline of all interior bents for a number of years. As acompromise, contractors proposed a plastic crack forming strip be placed in the wet concretecombined with continuous placement. This method, referred to as a controlled joint, wasused successfully on a number of projects and was made standard practice in the early1990s.

Dowels projecting from the substructure into the ends of the outside beams had been usedfor years on simple spans and were then used on some early continuous deck only projects.Although a slotted hole was specified, the ends of the beams often cracked severely. Theassumption was that the beam was often erected with the dowel at the back of the slot. Thiscaused the beam to bear directly on the dowel, and thermal shrinkage stresses would breakthe end of the beam. The remedy was to eliminate dowels entirely for the ends of units andto change the standard distance from centerline of bent to the dowels from 7 1/2 in. to 8 1/2in. Eventually tight round holes in the beams were eliminated entirely, and only slottedholes were specified at the interior bents where the slab was continuous.

Maintenance Issues. Maintenance problems have been confined to the bridges withinsufficient provision for expansion at the ends of long units. Beam ends cracked beforedowels were removed. There is always a slab crack in continuous slabs over the bent. Onseverely skewed structures where continuous placement was used without a controlled joint,the crack is more noticeable as it scallops across the roadway. The cracks are usually smalland have required little maintenance to date. Current reasoning is that this crack can be noworse than a leaky expansion joint and only slightly worse than a transverse shrinkage crackwithin the span. From beneath the structure, the crack is noticeable to casual observation in



the overhanging portion of the slab. It tends to be irregular in path and leaks to some degree.A slab construction joint or controlled joints over the bent and parallel thereto as used incurrent practice made this crack straight and less noticeable.

Current Status

Concrete bridge decks are used on most of Texas’ bridges. Designs are standardizedaccording to the AASHTO empirical moment method.

Most bridges are designed with a cast-in-place (CIP) slab, with prestressed concrete panels(PCPs) as the contractor’s option. Extensive formwork is required for CIP slabs, makingtheir construction more difficult in some situations. Truss bars are no longer used forreinforcing CIP slabs.

Prestressed concrete panels are the preferred method of constructing decks on prestressedconcrete I-beams and U-beams and are used occasionally on steel beams and girders.However, PCP decks are not allowed under the following circumstances:

♦ Curved steel girder bridges

♦ Bridge widenings, at the future beam span adjacent to the existing structure.

♦ Stage construction, at the future beam span adjacent to the stage construction joint.

Deck continuity over interior bents is often used for simple span structures to reduce thenumber of joints. Two, three, and sometimes four spans are strung together to form units.

All of the advancements in design, materials, and construction appear to have solved themost critical concrete bridge deck problems, but efforts are still underway to improve thedurability of bridge decks.


All current Texas bridge standards use an 8 in. thick slab. For constructibility, an 8 in. thickslab should be used for the majority of the decks in Texas. The use of 7 1/2 in. or 7 3/4 in.thick slabs may be justified only if the span can be designed with fewer beams as the result.

To properly support the bridge railing, an 8 in. thick (minimum) slab overhang shall be usedregardless of the slab thickness between beams. Slab overhangs are designed according tothe AASHTO Specification with the further provision that a grid and plate analysis3 orfinite element analysis may be used to evaluate design moments. It is preferable that thewidth of slab overhangs be less than, or equal to, one half of the beam spacing.

Lap splices shall be designed according to AASHTO Specification requirements with thefollowing provision:

Required lap splice lengths shall be increased by a factor of 1.5 when using epoxy-coated reinforcing for all reinforcing steel in the slab.



A minimum 1 percent reinforcement ratio (longitudinally) shall be used in the slab at thenegative moment region of continuous steel units. A #5 (Grade 60) bar placed between eachbar T (temperature steel) in the top layer is usually sufficient. The additional steel shouldrun the length of the negative moment region and be developed into the positive momentregions.

A thickened slab end shall be used for prestressed beam spans, as standard practice with theconventional type end diaphragm as the option. Thickened slab ends are often used withsteel stringers as well (see TxDOT Bridge Detailing Manual). Thickened slab end standardsdetails are as follows for prestressed I-beams and prestressed U-beams, respectively.

♦ IBTS standard sheet: Thickened Slab End Details for Prestressed Concrete I-Beams

♦ UBMS standard sheet: Miscellaneous Slab Details for Prestressed Concrete U-Beams

Additional recommendations for cast-in-place decks, prestressed concrete panel decks, anddeck continuity for simple spans are provided below.

Cast-In-Place (CIP)

The majority of the bridge decks in Texas shall use the TxDOT standard CIP slab design,which is shown in Figure 7-2. The design meets the AASHTO Specification requirementswith the following provisions:

♦ Service load design is used for conservatism.

♦ Calculated stress (fs) in the transverse reinforcing steel shall not exceed 24,000 psi.

♦ Calculated concrete stress (fc) shall not exceed 1,600 psi.

♦ The modular ratio (n) is taken as 8.

All bridge deck designs must meet the AASHTO requirements with the above provisions.No change in the design of slabs is made when permanent metal deck forms (PMDF) areused.

The TxDOT standard CIP slab design may be used without further analysis except underhighly unusual conditions. Such conditions may include the following:

♦ Clear span beam spacings used are greater than that specified by the TxDOT standardcast-in-place slab design.

♦ Widths of slab overhangs used are greater than that specified by the TxDOT standardcast-in-place slab design.

♦ Live loads are greater than HS20.

Unusual conditions such as these, and/or others, may warrant the use of a more detailedanalysis and design.



Figure 7-2. TxDOT Standard Cast-in-Place Slab Design (See following explanatory notes.Online users can click here to view this illustration in PDF.)


To promote consistency, TxDOT has developed a standard cast-in-place slab design.Design guidelines are as follows:

♦ Use #5 straight bars at 6 in. top (bars A) and bottom (bars B) for all slab thicknesses.

♦ Use #4 bars at 9 in. (maximum) for temperature steel in top layer (bars T).

♦ Use #5 bars at 9 in. (maximum) for distribution steel in bottom layer (bars D).

This design may be used without further analysis for the majority of bridge slab designs inTexas if the design provisions below are followed:

♦ Concrete strength f 'c = 4,000 psi

♦ Use Grade 60 steel for all reinforcing.

♦ Normal maximum beam spacing should be limited to 9 ft. Spacing may be increased to9.5 ft. for flared beam situations.

♦ Criteria in table below shall not be exceeded:

TxDOT Standard Cast-in-Place CriteriaSlab Thickness * Max. Clear Span

Beam Flangeto Beam Flange

* Max. OverhandBeam Flange to Face of

Rail at End of Unit

* Max. OverhandBeam Flange to Face

Rail at Midspan7.5 in. 7.896 ft. NA NA

7.75 in. 8.295 ft. NA NA8 in. 8.686 ft. 1.75 ft. 4.00 ft.

* Values shown are for concrete beams. Adjust to 0.25 point of flange for steel beams.



Prestressed Concrete Panels (PCP)

The TxDOT standard detail sheets for prestressed concrete panels on prestressed concretebeams and structural steel beams are available for insertion into the bridge plans only whenprestressed concrete panels are allowed on the project.

The TxDOT standard detail sheets for prestressed concrete panels are as follows:

♦ PCP (C) standard sheet, for use on prestressed concrete I-beam spans.

♦ PCP (U) standard sheet, for use on prestressed concrete U-beam spans.

♦ PCP (S) standard sheet, for use on structural steel beams or girders.

The designs represented on these details are highly standardized, but they are intended tofollow AASHTO with the following provisions:

♦ Service load design is used but ultimate strength is checked at mid-span.

♦ The panel alone supports dead load. The composite PCP/CIP slab prestressed crosssection resists live load positive moments. Transverse reinforcing in the CIP portion isdesigned for the negative live load moment.

♦ Distribution reinforcing is not required. This is justified by successful load testing.

♦ Calculated stress (fs) in the transverse reinforcing steel does not exceed 24,000 psi.

♦ Calculated tensile stress in the bottom of panel does not exceed c' f6(where f 'c = 5,000 psi, ft = 424 psi).

♦ Ultimate flexural capacity at mid-span is based on the following:

32f

2D LengthPanelf se

su +=

Using 0.375 in. strands and( )( ) 452700.7fse −= = 144 ksi

♦ Design is based on the use of 3/8 in., 270 kip strand, stressed to 16.1 kip.

♦ Reinforcing steel #4 (Grade 60) at 6 in. spacing may be substituted for strands in panels5 ft.-0 in. width or less and shall be required in panels 3.5 ft. width or less.The problem here is that panels are generally not wide enough to develop larger strands,so the design is based on the amount that can be developed, rather than the fulldevelopment. Very short panels must have mild steel reinforcing instead of prestressingstrand to prevent splitting.



All designs must meet these requirements and should conform to the TxDOT standard detailsheets. The TxDOT standard detail sheets are to be used without further analysis exceptunder highly unusual conditions. Such conditions may include the following:

♦ Clear span beam spacings used are greater than that specified by the TxDOT standardcast-in-place slab design.

♦ Widths of slab overhangs used are greater than that specified by the TxDOT standardcast-in-place slab design.

♦ Live loads are greater than HS20.

Unusual conditions such as these, and/or others, may warrant the use of a more detailedanalysis and design.

Deck Continuity for Simple Spans. Rather than giving design controls, which would befactitious for this type of construction, recommendation will be made regarding details thatappear to function acceptably:

♦ The total length of continuous deck slab is controlled by the capacity of the joint and bythe ability of the end bearing to move without slip. Although some units have beendesigned to upward of 400 ft. in length, this is not common practice and can rarely bejustified with standard bearings.

♦ The capacity of the standard bearings as detailed on the IBB sheet must be checkedaccording to the latest design procedures. (See Bearings in Chapter 9, Section 11.) Theslip capacity of the bearing is significantly controlled by the dead load reaction of thebeam in the end span. Therefore, long end spans are beneficial to the performance ofthe unit.

♦ A deck expansion joint should be provided at abutments or no more than one bentremoved from the abutments. Standardized systems such as open armor joints, sealedarmor joints, or 4 in. sealed expansion joints have adequate capacity for short ormoderate length structures with two expansion joints. However, long bridges havemultiple units that expand into each other, and the capacity of the joint becomes asignificant factor in selecting joint location and, therefore, unit lengths.

♦ Slabs should be detailed with construction joints or optional controlled joints at thebents. Controlled joints, which use a plastic crack former, are detailed on the IBMSstandard. Under no circumstances should the contractor be allowed to saw the jointafter the concrete has set as a substitute for the other methods. The random crackingmay not be apparent at first, but it will show up later; a saw cut will intercept therandom crack, increasing the possibility of spalling in this area.

♦ At each interior bent, a chamfer line should be required in the bottom of theoverhanging slab. The chamfer line should begin at the outside face of the top flange, orthe end of the inverted tee cap, and extend along the bentline to a line at a 2 ft. paralleloffset then extend normally to the edge and vertically up the outside face of the slab.See TxDOT Standard IBMS.

This will usually contain the crack and render it less conspicuous from below.



♦ At bents where adjacent spans have different beam spacing, bottom longitudinal bars,from the span with the smallest beam spacing, should be extended a minimum of 2 ft.into the other span. Top longitudinal bars should be spaced the same in adjacent spansand simply continue across the joint.

♦ Thickened slab ends must be used at expansion joints (see Figure 7-20). No slabthickening is required at the interior bents where the slab is continuously reinforced.

♦ Dowels projecting from the substructure into slots in the beam ends should be omittedat ends of units in most cases. There have been a few bridges built on extreme gradeswhere dowels were used at the ends of short units.

All other details and the design of the deck slab and beams should be the same as for simplespans.

Chapter 7 — Superstructure Design Section 2 — Two-Way Deck Slabs on Stringers


Section 2 Two-Way Deck Slabs on Stringers

Background

This slab design was not used early and has not been used often for Texas’ bridges. The firstknown use was in 1967 on a steel plate girder unit across the Trinity River. Plans wereunderway to make the river navigable as far up as Ft. Worth, and it became impossible for awhile to get a permit for a bridge without navigational clearances. This resulted in severalsuch bridges being constructed with a channel span in excess of 300 ft. It was found thatminimum steel weight and, consequently, maximum economy could be achieved by usingfewer girders and no stringers. To minimize slab thickness, the AASHTO provisions of“Slabs Supported on Four Sides” were applied.

There have been less than 20 bridges constructed with this type of deck in Texas.Experience has been favorable except for a few construction problems. One problem wasexcess slab cracking where the positive moment shear connectors stopped. This occurrencewas attributed to shrinkage in the long negative moment placement. Another problemoccurred on a unit that had shear connectors in the center span and none in the end spans.After placement, the slab was observed to lift off the top of the girder flange by as much as0.02 ft. This was also attributed to shrinkage, and the slab later crept down to bear on theflange.

Current Status

The two-way slab is a viable design for spans in which girder spacing in excess of 15 ft. iseconomical.


Two-way slabs must bear longitudinally on the girder and transversely on the floor beams. Ifgirders and floor beams are spaced equally, the wheel load is shared equally by the twospans. If one span is three times the other, virtually all of the load is carried by the shortspan. A ratio above 1.5 will result in very little savings in slab depth over a one-way slab.

A grid and plate computer program4 finite element analysis, or influence surface charts5

could be used for analysis, but it will usually suffice to design as follows:

♦ Calculate the percentage of load carried by each span according to the AASHTOformulas.

♦ Calculate transverse live load moments by the AASHTO empirical method. Multiply bythe appropriate span factor.

♦ Calculate longitudinal moments according to influence lines for continuity and theAASHTO distribution formula for slabs reinforced parallel to traffic. Multiply by theappropriate span factor.

Chapter 7 — Superstructure Design Section 2 — Two-Way Deck Slabs on Stringers


♦ Specification requirements for distribution steel may be ignored if the span ratio is 1.5or less.

♦ In negative moment regions of continuous steel units, a minimum 1 percent oflongitudinal slab steel should be provided.

♦ Calculate overhang moments according to AASHTO.

♦ Select slab depth and reinforcing based on fc = 1,600 psi and fs = 24,000 psi.

Chapter 7 — Superstructure Design Section 3 — Simple Concrete Slab Spans


Section 3 Simple Concrete Slab Spans

Background

This is one of the earliest superstructure types designed to cross Texas creeks andbackwaters. Hydraulically, slab spans are better than box culverts. Economically, however,they have never been desirable. Substructure costs are excessive due to the limited distancebetween supports. With open railing, slab spans can become a nice solution for lowheadroom stream crossings where occasional flood inundation is expected.

AASHTO Specifications for slab spans have changed over the years, but not so drasticallyas for slabs on stringers. Deterioration has not been a significant problem because slabspans are thicker and are located predominantly in rural areas which are not salted.

History

In 1944, slab spans were being constructed according to the third generation of BridgeDivision standard details. Creation of a “farm-to-market” highway system was imminent,and the Highway Department wanted to build these roads as economically as possible. TheBridge Division became interested in experimental work being conducted on slab spans atthe University of Illinois. A design procedure was reported in their Bulletin 346 thatrecognized the contribution of the curbs in carrying load and, consequently, resulted inthinner slabs being required. Testing had been on models only, so a full-size slab bridgewas constructed in Henderson County and load tested to verify the Illinois method. Thetests revealed no significant problem and a new series of standard details, called FS Slabs,were developed around the concept. Design loads were H10 and H15, curbs were 18 in.high, no railing was used, and the slabs were unusually thin. Many bridges were constructedon the farm-to-market system according to these standards and have performed well underheavier than design loads. Unfortunately, many of them have required widening because oftraffic. Current policy requires widened structures to support at least H20 design load atservice load stresses. When the curbs are removed, the thin slab will no longer satisfy theoriginal design loads when analyzed by ordinary procedures. By strengthening the brokenslab edge and analyzing with a grid and plate computer program, it has been possible towiden the H15 designs. Widening of H10 designs has been done but is impractical.

Around 1948, a new series of slab spans for all highways was developed. Concession wasmade to constructibility by adjusting span lengths for different skews so that the same lengthof bar joist could be used to support the forms. Nominal 25 ft. slab spans were 25 ft. centerto center of bent for 0 and 15 degree skew, 25.5 ft. for 30 degree skew, and 26.25 ft. for 45degree skew. The design span length and main reinforcing steel was parallel to traffic up to15 degree skew but perpendicular to the bents for 30 and 45 degree skews. The lattermethod saves concrete but causes problems when widening is required. These standards areno longer used.



Design Issues

There have been two significant performance problems observed with concrete slab spans.Skewed spans tend to work themselves out of line laterally because of closed expansionjoints over a smooth bearing surface. This tendency can be prevented with a substantialshear key in the bearing surface. Where the width of a slab span, along the centerline ofbent, exceeds about 40 ft., shrinkage cracking perpendicular to the bent can be alarming.This is especially evident at fixed ends, which are restrained from movement by dowels intothe cap. It can be minimized by placing all of the dowels in the center portion of the cap andlubricating the outside portions.

A few structures have been constructed with simple slabs precast in about 8 ft. widths, butthis is not encouraged.

Current Status

Most new simple concrete slab bridges currently being constructed are project specificdesigns conforming to Bridge Design Section or district design practice.

Precast reinforced concrete slabs are discouraged. Widening of FS slabs is discouraged, andis prohibited for H10 designs.


For new simple concrete slab span designs, the following practice is recommended:

♦ Class S concrete is required.

♦ For skews through 30 degrees, place the main reinforcing and longitudinal temperaturesteel parallel to traffic. Use design span length parallel to roadway for skews up to 15degrees. Use design span lengths perpendicular to bents for skews over 15 degrees, butdivide the required area of main reinforcing by the square of the cosine of the skewangle. Place transverse reinforcement parallel to the skew.

♦ For skews over 30 degrees, place the main reinforcing and longitudinal temperaturereinforcing perpendicular to the bents and calculate effective span length accordingly,between quarter points of the bent cap width. Place transverse reinforcing parallel tothe skew.

♦ Calculate wheel load distribution as 4 + .06 x Effective Span for both cases.

♦ Use live load moments from the AASHTO Specification6 or calculated from militaryloading.

♦ Service load analysis allowing fc = 1,600 psi in concrete and fs = 24,000 psi in steel isrecommended. However, load factor design is also acceptable.

♦ Provide distribution reinforcing according to the AASHTO Specification.

♦ Effective span should be no more than 20 times total slab thickness to control long-termdead load deflection.



♦ The edge beam live load moment of 0.1 PS may be applied to a slab width of 4 ft. Thishas been the practice and has shown to be acceptable through experience. For skewsthrough 30 degrees, the reinforcing required by the primary slab design will usually besufficient for the edge beam. For skews over 30 degrees (main reinforcingperpendicular to the bents), separate reinforcing parallel to the edge must be provided.

♦ Check embedment length of bottom reinforcement steel, across the face of the cap at theends of units. Hooked bars may be required.

Typical bearing and joint details are shown in Figure 7-3.

Figure 7-3. Reinforced Concrete Slab Spans – Typical Details (Online users can click hereto view this illustration in PDF.)

Chapter 7 — Superstructure Design Section 4 — Continuous Concrete Slab Spans


Section 4 Continuous Concrete Slab Spans

Background

The use of continuous concrete slab units began around 1936. Spans were from 20 to 30 ft.Some of the early designs required bent caps to be placed monolithically with the slab, apractice that was later discouraged because of construction difficulties. Later designsextended the cap stirrups around large transverse bars in the slab, effectively deepening thecap for live load.

Interior spans up to 40 ft. were occasionally constructed with constant depth slabs.

The use of variable depth slabs began in the 1950s. Interior spans up to 60 ft. were feasible.Many were constructed on interstate and primary main lanes. One district designed a fewinterstate crossover structures as slabs with a parabolic soffit. These were attractive bridgeswith open concrete railing.

The longer spans became unpopular because of long-term deflections. If the riding surfacewas constructed according to calculated short-term deflections, it had to be overlaideventually to restore the riding quality. If cambered to account for long-term deflections,the bridge would give a rough ride during its early life. Neither situation was considereddesirable.

Another objection to continuous slabs was the amount of cracking that occurred in thenegative moment area, especially when curbs and concrete railing were provided. Curbswere usually not considered in the design but they try to help anyway. Of course, reinforcedconcrete is supposed to crack in a tensile zone, but nobody seems to like it when it actuallyhappens. Creep in the concrete, which caused the deflection problem, also aggravated thecracking.

Current Status

Variable depth reinforced concrete slabs are no longer recommended. Constant depthreinforced concrete slabs are a logical solution for many stream crossing problems, but theyare relatively expensive. Because of fewer deck joints, continuous units are often preferredover simple spans. Continuous unit lengths to 150 ft. are used with the ends connectedsecurely to the abutments. Unit lengths of 200 ft. may be used with both ends free toexpand.


For new continuous concrete slab span designs, the following practice is recommended:

♦ Do not design continuous reinforced concrete slabs for skews over 30 degrees.



♦ For constructibility, it is desirable that all span lengths (face to face of bent) be equal forbridges of two spans and bridges of four spans or greater. Bridges of three spans shallbe proportioned in 1.0-1.2-1.0 span lengths to balance the unit.

♦ Place the longitudinal reinforcing parallel to traffic. Use design span lengths parallel totraffic. Place transverse reinforcing parallel to skew.

♦ Effective supports may be taken at the centerline of interior bents and at the quarterpoint of end bent caps.

♦ Calculate wheel load distribution as 4 + .06 x Effective Span (7.0 ft. maximum). Use ofinterior span length is usually close enough for end spans also.

♦ Composite moments and reactions may be calculated from constant moment of inertiainfluence lines, or computer programs, assuming knife edge supports. Truck loadingwill control both positive moment and negative moment for span lengths below 30 ft.Military loading must be investigated as well.

♦ Design negative moment may be taken at the face of bent cap.

♦ Service load analysis allowing fc = 1,600 psi in concrete and fs = 24,000 psi in steel isrecommended. However, Load Factor design is also acceptable.

♦ Provide distribution reinforcing according to the AASHTO Specification.

♦ Effective span should be no more than 23 times total slab thickness to control long-termdead load deflection. This has been the practice and has shown to be acceptable throughexperience.

♦ Check embedment length of bottom reinforcement steel, across the face of the cap at theends of units. Hooked bars may be required.

Typical bearing and joint conditions are shown on Figure 7.4.



Figure 7-4. Continuous Reinforced Concrete Slab Units – Typical Details (Online users canclick here to view this illustration in PDF.)

Chapter 7 — Superstructure Design Section 5 — Simple Concrete Girder Spans


Section 5 Simple Concrete Girder Spans

Background

Concrete girder spans have been constructed for Texas highway bridges since the creation ofthe Department. The earliest standard drawings were adapted from Bureau of Public Roadsdesigns in 1918. Several series of standard drawings followed until the last issue for the endspans of interstate crossovers in 1956.

For many years, cast-in-place concrete girder spans were more popular than steel I-beamspans. Many bridges were thus constructed in the 1920s and 1930s. Clear span lengthsranged from 16 ft. to 40 ft. with the majority being 26 ft. Most of these have now beenwidened because their roadway widths were too restrictive for the safety of modern traffic.

In the 1940s the use of concrete girders faded in favor of steel I-beam spans. Precastreinforced concrete girders were used on a few projects to widen existing cast-in-placeconcrete girder spans.

Design parameters are tabulated in Figure 7-3. Typical standard details are shown inFigure 7-5, Figure 7-6, and Figure 7-7.

Construction Issues

Construction problems were primarily due to the complication of forms and falsework. Inthe earlier years, forms were supported by falsework resting on the ground throughout thespan. Later, contractors began to use steel joists, which could span from bent to bent for theaverage span. Concrete had six sacks of cement per cubic yard and no air entrainmentagents, retarders, or super water reducers. Concrete was mixed on the job in batchesconsiderably smaller than one cubic yard. Many of these bridges exhibit excellent concretework and have been very durable.

Maintenance Issues

Very little deck maintenance has been required on the old concrete girders. When deckdeterioration became serious on Texas bridges, one district engineer, who was conductinghis own investigation to determine the cause, enjoyed demonstrating the durability of oldconcrete girders. There were several locations in the district where 1930 vintage concretegirders had been paralleled by pan form girders cast in the 1950s using five-sack concretewith air entrainment. Deterioration had begun on the newer bridge, while the old bridge wasstill sound. He was arguing for a return to six-sack concrete without chemical admixtures.Current specifications require six sacks of cement per cubic yard, but air entrainment is stillused.

The roadway surface of concrete girder spans tended to be rough because of excessdeflections due to creep of the concrete. This problem could usually be remedied by anasphalt overlay.



Malfunctioning bearings caused numerous maintenance problems on concrete girders.Frozen bearings often caused girder ends to break off. The solution was usually to transferthe reaction to a pedestal constructed under the end diaphragms.

Current Status

Simple concrete girder spans are no longer used.


If there should be an appropriate occasion to use simple concrete girders, the followingrecommendations are offered:

♦ Limit spans to 45 ft. with a maximum span-to-depth ratio of 14.

♦ Distribute live load laterally as specified for concrete deck on concrete T-beams.

♦ Use service load design, Grade 60 reinforcing, and 4,000 psi concrete.

♦ Shear stress carried by concrete alone should be taken as 60 psi.

♦ Long-term deflections should be calculated for the gross section assuming a modulus ofelasticity of 1,500,000 psi. This value is based on the creep of the concrete. Experienceand observation has shown this value to be adequate.

Table 7.3: Chronology of Concrete Girder StandardsAllowable Stress (psi)

Series Year Designed Live Load Steel Concrete1G1. . . 1918 20T Roller 16,000 650

1G263. . . 1920 Typ. 15T Truck 16,000 650DG-5. . . 1928 2-15T Trucks 16,000 650

G-24-28.5. . . 1933 2-15T Trucks 16,000 650GL-22. . . 1937 H10 16,000 650G-2-24. . . 1938 2-15T Trucks 16,000 900G-28H. . . 1951 H20 20,000 1,000

G-28HS. . . 1951 H20S16 20,000 1,0002G-26(4)-35. . . 1956 H20 20,000 1,200

1. Adapted from Bureau of Public Roads and Rural Engineering Designs.2. For Interstate Highway Crossovers.



Figure 7-5. Evolution of Concrete Girder Standards – 1918, 1920, and 1928 (Online userscan click here to view this illustration in PDF.)



Figure 7-6. Figure Evolution of Concrete Girder Standards – 1933, 1937, and 1938(Online users can click here to view this illustration in PDF.)



Figure 7-7. Evolution of Concrete Girder Standards – 1951, 1956, and 1956 (Online userscan click here to view this illustration in PDF.)

Chapter 7 — Superstructure DesignSection 6 — Concrete Pan Form Slab and Girder

Spans


Section 6 Concrete Pan Form Slab and Girder Spans

Background

Pan form girders were developed in the late 1940s in anticipation of a need for low-costbridges on a farm-to-market road system soon to be funded. The terminology depicts themodular steel forms required for the cast-in-place reinforced concrete spans. Whenassembled, bolted together, and supported from the bent caps, a metal pan was provided toform the concrete and support the weight in flexure without intermediate support. Formsand falsework are combined in a sturdy reusable package. The original span length was 30ft. for 20 in. wide caps and no skew. It was soon discovered that trestle piling would seldomfit inside a 20 in. wide cap. Cap width was changed to 24 in. and, since the distance face toface of caps had to remain the same to allow form removal, the basic span length became 30ft.-4 in.

In 1956 a design was introduced for 40 ft. span, and someone discovered the possibility ofaccommodating skew by offsetting adjacent form barrels. With the faying joint surfaces 36in. apart at the center of girders and bolts on 9 in. centers, an offset of one bolt spacingproduces a skew angle whose tangent is 0.25. This accounts for the unusual skew anglescovered by the standard details. Each skew increased the span length by 10 inches: 9 in. forthe bolt spacing and an extra inch to facilitate form removal. In the early 1960s, standarddrawings were distributed for superstructure and substructure for both span ranges, fiveroadway widths, and the five skews. Many bridges have been constructed to thosestandards, as well as those previous and others that followed.

Pan form girders have been the most economical method for constructing a highway bridgefor most of their existence. In the last few years, however, prestressed concrete beams haveovertaken them. Standard details have usually been available for pan form girder spans andsubstructure to fit the geometric requirements of most stream crossings. This allows thedistrict to prepare a complete set of plans for a project with only a foundation review by theBridge Design Section. In spite of their maintenance problems, approximately 3,688 panform girder bridges had been constructed on the Texas highway systems by 1999.

Pan form spans were more of a development than a design. The interesting part was thedevelopment and design of the forms themselves. Once standardized, it is impossible tochange the depth or spacing of girders without retiring considerable contractor investment informs. Other span lengths are not considered appropriate. Longer spans would haveexcessive long-term deflection and shorter spans can be handled more effectively withconcrete slab spans.

The evolution of pan form girder standard details is depicted in Figure 7-8, Figure 7-9, andFigure 7-10.


Spans


Figure 7-8. Evolution of Concrete Pan Form Slab and Girder Standards – 1948, 1950, and1952 (Online users can click here to view this illustration in PDF.)


Spans


Figure 7-9. Evolution of Concrete Pan Form Slab and Girder Standards, 1956, 1959, and1964 (Online users can click here to view this illustration in PDF.)


Spans


Figure 7-10. Evolution of Concrete Pan Form Slab and Girder Standards – 1966, 1991, and2001 (Online users can click here to view this illustration in PDF.)

The design was created by ingenuity and input from contractors. The AASHTOSpecification did not exactly anticipate this type of structure. Designs in the 1960s appear tohave deducted the flexure and shear strengths of the diaphragms from the calculatedmoments and shears in the cap. Design and detailing became onerous when numerouscombinations of span, roadway, and skew with two different types of foundation had to beproduced for standard series.


Spans


The concrete thickness at the crown between girders was 3.5 in. thick in the beginning. Inthe early 1960s small longitudinal cracks were noticed through this thin section for most ofthe length of some of the outside barrels. In spite of Texas insistence that the cracks wereharmless, the FHWA insisted that the section be thickened to 4.5 in. This caused a stiramong contractors who owned forms that had to be modified. The cracking still occurs inthe thicker section but everyone is accustomed to it now.

Model tests and field studies7 in 1969 indicated the distribution of wheel loads to eachgirder to be about 25 percent less severe than assumed in the design. Shear was not asignificant problem. Also the large end diaphragms doubled the strength of the caps againstlive load. No adjustments were made to the design except to convert the flexural reinforcingin the girders to Grade 60 at the ratio of 20/24 times Grade 40 area.

Construction Problems

Construction problems were worked out through experience. Outside girder forms werehard to hold in line, especially when the 1960 standards required some deck slab overhangpast the outside girder face. Over the years widths of slab overhangs have varied fromnothing in the beginning (Figure 7-8) to a maximum of 1 ft.-9.25 in., which was difficult toconstruct (Figure 7-10). In the early 1990s the slab overhangs were reduced to a maximumof 7.75 in. due to concerns about railing anchorage. Later, in the 2001 standard design detailsheets, the slab overhang depth was increased from 6 in. to 8 in., and the slab overhangwidened to a maximum of 1 ft.-1.75 in. When skews were introduced, the bent caps wereallowed to extend past the outside girder face. This trapped the outside form so that it had tobe cut out around the bent cap to allow removal. The greater the skew, the further the centerof form support was from the face of the cap. Careful attention had to be given to the formsupport member and its connection to the cap to prevent rotation under load. Trestle pilebents were limited in height because of the eccentric load caused by concrete placement onthe side of the bent. Deflection of the forms due to the weight of concrete was significant.The forms could be cambered to counteract this, but the camber usually came out afterseveral uses, which caused extra depth of the girders and a sag in the bottom of the stem.Long-term deflection of the girders was also a factor. If the surface was cambered toaccount for this, the bridge gave a rough ride when it was opened to traffic. If not, it becamerough after three years. Neither was desirable.

Maintenance Problems

Unfortunately, maintenance problems have been quite extensive for pan form bridges. Themajor problem occurs when dirt enters the joint over each bent and cannot get out becauseof the cap underneath. This occurrence allows the joint to open but not to close so that, in afew years, the joint growth has shoved the spans outward across the cap. If the spans arefixed to the cap with reinforcing bar dowels, large slivers are broken from the cap sides.Even without dowels, the movement can eventually damage the cap. Attempts have beenmade to alleviate this problem by eliminating fixity dowels on both sides of the cap,reinforcing fixed deck joints across the top, and providing an opening beneath expansionjoints for dirt to fall through. The latest details are shown in Figure 7-11. These measures


Spans


have been successful to some degree, but there were already many pan form bridges withadvanced symptoms before the corrective measures went into effect. There have been a fewbridges with spans shoved almost off the cap and spans imbedded into the approachembankment because of joint growth. Repair of the advanced cases is tedious.

Figure 7-11. Pan Form Slab and Girder Joint Details (Online users can click here to viewthis illustration in PDF.)

Current Status

Pan form girder spans will continue to be popular for constructing stream crossing bridges.The FHWA discourages pan form construction on interstate highways. Standard details


Spans


were revised in 1991 to drastically reduce the amount of slab overhang past the outsidegirder face. Research has indicated that the required bridge railing strength cannot bedeveloped in the 6 in. thick slab overhang that previously was constructed.

In 2001, slab overhangs were increased in thickness to 8 in. and limited in width to 1 ft.-1 3/4 in., to maintain the strength of the railing connection.

Standard design detail sheets that cover five roadway widths and contain all of the designimprovements previously discussed are available on the TxDOT web site.

It appears that pan form bridges will continue to be a significant part of TxDOTconstruction.


The new standards are designed, basically, according to the current AASHTO Specification.The old methods and approximations have been abandoned. Any nonstandard span shoulduse the same girder design. Salient features of the design are as follows:

♦ Concrete should be Class S, currently 4,000 psi 28-day strength, and reinforcing steelshould be Grade 60.

♦ Service load design should be used.

♦ Distribution of wheel loads should be taken as S/6.0 according to the original design.

♦ Stirrup design should comply with the latest AASHTO Specification using 60 psi shearstress carried by the concrete.

♦ 8 in. thick deck slab overhangs should be limited in width to 1 ft.-1 3/4 in. This willprovide adequate railing support and alleviate the construction problems associated witha wide overhang.

♦ Conventional bent design procedures are recommended.• Live load to deck as two 20.8k concentrated loads and 10 ft. uniform load• CAP 18 analysis• No allowance for diaphragm strength• Load factor design• fy = 60 ksi• νu = 126 psi• Service load reinforcing steel stress controlled by crack width (Z = 170)• Dead load reinforcing steel stress < 22 ksi

Chapter 7 — Superstructure Design Section 7 — Continuous Concrete Girder Spans


Section 7 Continuous Concrete Girder Spans

Background

Except for a few early special structures, Texas’ use of continuous reinforced concretegirders began in the 1950s and ended in the 1960s. This type of structure, with parabolicgirder soffits, was considered highly aesthetic by a few engineers.

The Waco District constructed the majority of these bridges, using cast-in-place concretesuperstructures for many bridges on the interstate and primary highway systems in thedistrict. Possibly because of construction volume, cast-in-place bridges were fairlyeconomical.

This economy was not as evident in other areas, where concrete girder units were usedprimarily to satisfy aesthetic opinions. Some of these were constructed on major highwaysin Austin, Amarillo, and Wichita Falls, and over interstate highways in the Abilene andBryan Districts.

The Bridge Design Section prepared two sets of standard designs for interstate crossovers in1956. These were used for only the few concrete girder bridges in the Abilene and BryanDistricts. All such bridges in the Waco District were designed in the district bridge office.Wichita Falls’ bridges were special designs by the Bridge Design Section. Austin andAmarillo variable-depth concrete girder units were designed and detailed by one meticulousengineer who worked in both districts and the Bridge Design Section at various times.

A variable-depth cast-in-place continuous reinforced concrete girder bridge, constructed inHill County in the early 1960s was used for significant research on high-strength reinforcingsteel.8 . The bridge had three girders with spans of 50-88-88-50 ft. on 30 degree skew. Itwas load factor designed to H-15 loading using 3,000 psi concrete and Grade 60 reinforcing.Actual concrete strengths exceeded 5,000 psi. The research pronounced the performance ofthis structure completely satisfactory.

This type of structure was not used after 1968 because of the complication of falsework andforms, which increased costs and usually increased construction time. Forming over trafficis especially hazardous.

Maintenance Issues

Maintenance problems have been few. Long-term deflections are significant, but if initialcamber is acceptable this effect can be controlled. Asphaltic concrete overlay can be appliedif deflections become excessive. The structures are resistant to impact from overheightloads because of their mass. The only serious problems with concrete girders have occurredwhere de-icing salts are used extensively. These structures, built in the 1950s, havedeteriorated severely. There appears to be no way to neutralize this problem other thancomplete removal and replacement of the bridge.

Chapter 7 — Superstructure Design Section 7 — Continuous Concrete Girder Spans


Current Status

Continuous reinforced concrete girders are no longer used.


If there should be an appropriate occasion to use continuous reinforced concrete girders, thedesign should closely follow the AASHTO Specifications with the following considerations:

♦ Concrete should be Class S, currently 4,000 psi 28-day strength, and reinforcing steelshould be Grade 60.

♦ Load factor design should be used.

♦ Lateral distribution of live load should be S/6.0.

♦ Stirrup spacing should be calculated assuming the stress carried by concrete alone to be126 psi.

♦ Service load stress should be based on crack control criteria(z = 130 for negative moment, Z = 170 for positive).

♦ Long-term deflections should be calculated for the gross section assuming a modulus ofelasticity of 1,500,000 psi. This value is based on the creep of the concrete.Experience and observation has shown this value to be adequate.

Chapter 7 — Superstructure Design Section 8 — Concrete Box Girder Spans


Section 8 Concrete Box Girder Spans

Background

In 1958 a continuous cast-in-place box girder bridge (80-95-80 ft.) was constructed on SH240 over the Wichita River in Wichita County. This bridge has since been replaced due toheavy chloride damage. In 1962 concrete box girders were used in a railway service roadstructure in Dallas. Shortly thereafter the Preston and Pearl Street Underpass wasconstructed over IH 20 with a 14 ft. wide spine beam containing three 3 ft. round voids andwith 9 ft. cantilevered deck slabs. All of these were mild steel reinforced.

In 1974 a controversial freeway project in San Antonio was continued with numerousaesthetic and environmental embellishments, one of which was a continuous reinforcedconcrete box girder bridge. It had nine spans, the longest of which was 93 ft. The sides weresloped, and there was a sharp horizontal curve at each end of the bridge. The cost was $25per square foot compared to prestressed beam bridges at $16.

Then, in 1979, a small box girder unit (86 to 105 ft.) was constructed across a depressedsection of Spur 366 in Dallas. The cost was $65 per square foot compared to $16 per squarefoot for prestressed beam bridges.

The high cost of these structures has resulted in very few continuous cast-in-place concretebox girder designs of the type shown in Figure 7-12.

Similar sloping web box girders have been used sparingly to finish off the outside of precastbox beam bridges to accommodate severely flared roadways.

Current Status

Cast-in-place mild steel reinforced concrete box girders are not recommended except forvery special conditions.


No recommendations are considered necessary.

Chapter 7 — Superstructure Design Section 8 — Concrete Box Girder Spans


Figure 7-12. Cast-in-Place Concrete Box Girder – Typical Section (Online users can clickhere to view this illustration in PDF.)

Chapter 7 — Superstructure Design Section 9 — Prestressed Concrete Deck Slabs


Section 9 Prestressed Concrete Deck Slabs

Background

For some time it has appeared desirable to prestress bridge deck slabs to counteractcorrosion problems associated with cracking and lack of durability of the concrete.Unfortunately, a satisfactory design has not been developed.

A large elevated highway project on IH 345 in Dallas, Texas, constructed in the late 1960s,had 10.5 in. post-tensioned slabs on steel plate girder spans. The slab rested on transversefloor beams a few in. above the girders. Approximately 27 acres of this type slab wereconstructed on this elevated section as part of an attempt at aesthetics. Because of a myriadof streets and utilities beneath the structure, support location was random. It was consideredimpractical and obtrusive to provide bent caps, so widely spaced girders were supporteddirectly on the columns. The random column spacing and wide girder spacing combined tocreate some unorthodox continuous units. Concern about the ability of a conventional slab toconform gracefully to the anticipated warping led to the slab being supported by floor beamsonly and post-tensioned in both directions. The benefits of composite action were sacrificed.Compared to adjacent sections of this same highway, the bridges appeared to cost about$3.50 per square foot extra. The average bid price for structural steel was $0.29 per pound,and the slab cost was about $4.00 per square foot. The prestressed slab is still in goodcondition but the design has not been used since.

A research project at the University of Texas was directed to the design and performance oftransversely prestressed bridge decks.9 Model testing was done and corrosion exposurespecimens were evaluated. The report pronounced this a viable type of bridge deck with lifecycle cost about 20 percent less than conventionally reinforced designs. Six 75 ft.prestressed concrete beam spans in an interstate highway river bridge in La Grange, Texas,were constructed with 80 ft. wide prestressed decks. Some were stressed only transverselyand others stressed both transversely and longitudinally. Construction of the bridge deckswas fraught with problems. Delivery of the coated hardware was slow, and forming forblockouts was tedious. Strands were sensitive to misplacement due to foot traffic and wetconcrete loads. Stressing was time consuming. Unfamiliarity of the contractor and statepersonnel with the system probably magnified the inherent problems. There was somediagonal cracking in the outer portions of the deck width adjacent to construction joints.This occurred in some of the conventionally reinforced bridge decks as well. The consensuscause was shrinkage in the wide slabs. Failure of the prestressing to close these cracks wasconsidered significant. It was concluded that the possible advantages of the system wereoutweighed by the problems encountered during deck construction.

Chapter 7 — Superstructure Design Section 9 — Prestressed Concrete Deck Slabs


Current Status

The Bridge Division does not expect prestressed concrete bridge decks to be used on Texashighway bridges in the foreseeable future except for very special situations.


No specific design recommendations will be given here. If needed, general designinformation can be found in the report Application of Transverse Prestressing to BridgeDecks.10

Chapter 7 — Superstructure Design Section 10 — Prestressed Continuous Slab Spans


Section 10 Prestressed Continuous Slab Spans

Background

Prestressing in Texas began in the mid 1950s and soon became accepted as an effectiveprocedure to increase concrete span lengths and control deflections. Emphasis was placedon precast pretensioned beams, and the cast-in-place post-tensioned slab was relegated tospecial conditions where structure depth was critical or aesthetics were especially desirable.

Various configurations of prestressed concrete slab spans have been used. Separate interiorbent caps are usually unnecessary since the slab itself can be strengthened transversely tospan between columns.

Most of the designs have had variable slab thickness created by a parabolic soffit each sideof the interior columns. This lightens the dead load and allows the absolute minimumsuperstructure depth.

Center span-to-depth ratios as high as 50 have been achieved.

Severely skewed bents have been used, but this tends to ruin the aesthetic quality ofvariable-depth slabs.

Where superstructure depth is critical close to the interior supports, constant-depth slabshave been used. In order to lighten dead load, round voids have been formed in the positivemoment zones using fiber tubes. Span-to-depth ratios as high as 40 have been achieved.

Constant-depth prestressed concrete slabs can also be designed without voids. Because ofthe weight, additional prestressing is required for stress control. Prestressing for zero deadload deflections is usually impractical.


Most of the problems with these slabs have occurred during construction. Fiber tube voidsare subject to crushing or floating during concrete placement. Concrete consolidationbeneath the voids is difficult, and it is virtually impossible to prevent longitudinal shrinkagecracking above and below the voids. Deteriorating tubes can create methane gas which canactually rupture the concrete. Stressing of post-tensioning tendons creates initial problemson all types of structures. Agreement between jack pressure and elongation is often beyondthe specification allowable of 5 percent due to a difference between calculated and actualfriction losses. In one embarrassing incident, a variable depth slab offset itself laterallyabout 1.5 ft. due to stressing across a severely skewed construction joint. Since then, anyconstruction joints in prestressed slabs have been normal to the tendons, or substantial shearkeys are provided.



Current Status

Continuous prestressed concrete slab units continue to be acceptable solutions to aesthetic orcritical depth problems; however, they are rarely used. Voided slabs are no longerrecommended.


The following items should be considered in the design of continuous prestressed concreteslab units.

♦ Skews greater than 45 degrees should be avoided.

♦ Primary prestressing tendons, longitudinal reinforcing bars, and design span lengthshould be parallel to traffic.

♦ Transverse reinforcing bars and prestressing tendons, if used, should be parallel to theskew.

♦ Effective supports may be taken at the centerline of interior bents and the quarter pointsof end bent caps.

♦ Trucks and lanes of live load may be distributed over 14 ft. of width for spans over 50ft. This should suffice for most units.

♦ Compute moments and reactions using a beam analysis computer program such asProgram B30.11 Assume knife edge supports continuous for the width of the designsection.

♦ Use negative moment at the centerline of interior bent for design.

♦ Design of longitudinal prestressing tendons should be based on low-relaxation sevenwire strand tendons with f 's = 270 ksi, jacking stress = 0.75 f 's. Losses may becalculated using anchor set = 0.625 in., friction coefficient = 0.25, and wobble factor =0.0002. The stress at anchorage after seating shall not exceed 0.70 f 's. Tendon stress atthe end of the seating loss zone need not be checked. Losses after seating may be takenas 33 ksi.

♦ Required final prestress force at the critical points should be shown on the plans. Thecontractor has the option of selecting the size of strands and tendons and the type ofanchorage.

♦ Anchorage zone reinforcing against bursting may be evaluated using CTR ResearchReport 208-3F,12 or other rational methods.

♦ Stresses due to primary and secondary effects of prestressing can be calculatedaccurately using computer program BMCOL51.13

♦ Ultimate flexural capacity should be checked at the maximum moment points only.Show the required values on the plans.

♦ Ignore shear in the longitudinal design of the spans.

♦ Transverse design at interior bents may use prestressing or mild steel reinforcing. Shearshould be considered.



♦ Transverse mild steel should be a minimum of #4 at 6 in. centers top and bottom.Longitudinal reinforcing should be a minimum of #4 at 12 in. centers. Distribution steelrequirements of the AASHTO Specification do not apply.

♦ Calculation of pay quantities of “Prestressing” should be shown on the plans to avoidconfusion. Rules for measurement are in the construction specification.

Typical bearing and joint conditions are shown in Figure 7-13.

Figure 7-13. Continuous Prestressed Concrete Slab Units – Typical Details (Online userscan click here to view this illustration in PDF.)

Chapter 7 — Superstructure DesignSection 11 — Prestressed Simple Slab and Girder

Spans


Section 11 Prestressed Simple Slab and Girder Spans

Background

The use of prestressed simple-span concrete slab and girder spans has been sparing butnotable.

One such installation formed the second prestressed bridge constructed on the Texashighway system. In 1954 the contract was awarded for a section of Dallas expressway thatincluded the Pine Street Overpass containing three 58 ft. post-tensioned simple concrete slaband girder spans. The contractor elected to cast the spans on the ground and lift them intoplace. Post-tensioning was accomplished with high-strength bars. The danger of improperprestressing procedures was emphasized when a stressed bar, kinked by collapse of ananchorage shim, was launched into the work area, barely missing a contractor’s employee.

The only other known use, which could be classified as prestressed simple slab and girderspans, occurred in 1959 on the Lavaca Bay Causeway. This two-mile-long bridge offeredthree superstructure alternates: pan form girders, prestressed concrete beam spans, andmonolithically cast pretensioned slab and beam spans. In a rare occurrence, pan forms lostthe bidding competition to the monolithic spans.

The monolithic spans were patterned after the Lake Pontchartrain Causeway in Louisiana,with one less girder and a slightly longer 60 ft. span. The contractor cast these spans inretractable metal forms near the bridge end. The spans were carried to the water by a gantry,loaded to a barge, transported to their position in the bridge, and erected by a large, speciallyfabricated crane mounted on the barge.

Three spans on the east end of the bridge around elevation 15 ft. were lost to HurricaneCarla before the bridge was opened to traffic. They were replaced with prestressed concretebeam spans.

Current Status

No designs of this type are anticipated.



Chapter 7 — Superstructure DesignSection 12 — Prestressed Pan Form Slab and

Girder Spans


Section 12 Prestressed Pan Form Slab and Girder Spans

Background

This concept never caught on, but needs mentioning for its historical significance.

Texas’ first prestressed highway span was post-tensioned pan form girders. It wasconstructed in 1952 on SH 60 across the San Bernard River between Austin and WhartonCounties. The regular bid was for 24 25-ft. slab spans and one 50-60-50 ft. continuousI-beam unit. The alternate bid was for 23 30-ft. regular pan form girder spans and one 60 ft.prestressed pan girder span. The alternate won. For the main span, two sets of 30 ft. panforms were erected with a falsebent between. Two coated, large-diameter strand tendonswere draped in each stem. An approximate 2 ft. space was provided beyond one end forstressing the tendons. After the concrete reached a strength of 4,000 psi, the tendons werestressed and the falsebent removed. The stressing space was covered with a cast-in-placeslab. The bridge still carries traffic.

Also, a continuous post-tensioned bridge was constructed in the Waco District using panform girders in the positive moment areas with solid slabs over the supports. The bridgelooks good but, because of construction problems, the system will probably not be usedagain.

Current Status

No designs of this type are anticipated.



Chapter 7 — Superstructure DesignSection 13 — Prestressed Continuous Concrete

Girder Spans


Section 13 Prestressed Continuous Concrete Girder Spans

Background

Cast-in-place T-shaped girders are not desirable for long continuous spans because of largecompressive stresses in the narrow girder stems at interior supports. For span lengths anddepths within practical limits of the compressive stress, mild steel reinforcing is usuallysufficient to resist tensile stresses. For prestressing to be advantageous, something else mustbe done to control compressive stresses over the support.

Continuous prestressed concrete girders were used efficiently in crossover structures over adepressed freeway in Houston. T-shaped girders were used in the positive moment regionsof continuous units, while the negative moment regions were the same depth but solidconcrete slab. This is shown as a ribbed slab in Figure 7-14. Post-tensioning was providedin the T-girder stems with additional tendons in the slab portion anchored in the spacesbetween adjacent stems. Span-to-depth ratios approaching 38 were achieved with spans upto 150 ft.

Current Status

No further designs of this type are anticipated.


Although this type of construction is not anticipated in Texas, the following items areoffered for consideration:

♦ Span-to-depth ratio should not exceed 38.

♦ Length of the solid slab portion must be sufficient to maintain compressive stresses inthe bottom of T-girders below 0.4 f 'c.

♦ Liberal stem widths should be provided to allow ample side cover to the post-tensioningducts after allowance for construction tolerances.

♦ Positive moment prestressing should be aligned with the center of stems and anchoredat the ends of unit. Additional tendons required for negative moment should beanchored at the end of the solid slab portion in the space between girder stems.

♦ Shear should be considered in the solid slab and in the T-girders.

Chapter 7 — Superstructure DesignSection 13 — Prestressed Continuous Concrete

Girder Spans


Figure 7-14. Ribbed Slab Concept (Online users can click here to view this illustration inPDF.)

Chapter 7 — Superstructure DesignSection 14 — Prestressed Cast-in-Place Box Girder

Spans


Section 14 Prestressed Cast-in-Place Box Girder Spans

Background

In 1973 a cast-in-place prestressed concrete box girder (four spans at 155 ft.) in Belton,Texas, was designed as an alternate bid against curved steel plate girders. The steel designwas selected by the contractor.

While California developed this design as its standard type structure, Texas has not beenable to make it an economically viable alternative. Complicated falsework, forming, andextensive on-site construction time are the downfalls of this system.

The high cost of these structures has resulted in a very few continuous prestressed concretebox girder designs of the type shown in Figure 7-15.

Figure 7-15. Prestressed Concrete Box Girder – Typical Section (Online users can clickhere to view this illustration in PDF.)

Current Status

Longitudinally prestressed cast-in-place box girders may have limited application.


The PTI Box Girder Manual14 and CALTRANS Design Practice15 are good references forthe design of cast-in-place prestressed concrete box girders. Numerous provisions of theAASHTO Segmental Guide Specification16 also apply to this type of construction.

Preliminary design should be very thorough. Decisions regarding depth of box, bottomwidth, number of webs, and type of deck reinforcing should be based on economic

Chapter 7 — Superstructure DesignSection 14 — Prestressed Cast-in-Place Box Girder

Spans


calculations. Web thickness should be sufficient to comfortably accommodate theprestressing tendons and withstand shear and torsion forces near the interior supports.

Final design should include allowances for creep and shrinkage, as they affect prestresslosses and redistribution of moments in continuous structures. Moment redistribution fromprogressive construction of continuous spans should also be considered.

Suggestions based on limited design experience are given below.

♦ Span-to-depth ratios should not exceed 20 for end spans nor 25 for interior spans.

♦ Prestressed concrete should be Class H with a maximum f 'c = 6,500 psi. Field controlby cylinder tests (Class H concrete) should be required.

♦ Anchorage zone reinforcing should be given careful attention.17 This was the generallyaccepted basis for consideration of lead-in stresses, but now an equilibrium-basedplasticity model (strut-and-tie) as described in the standard specification may be moreappropriate for sizing anchorage zone reinforcing.

Chapter 7 — Superstructure DesignSection 15 — Prestressed Segmental Box Girder

Spans


Section 15 Prestressed Segmental Box Girder Spans

Background

Concrete box girders can be built by connecting full cross-section cast-in-place or precastsegments with longitudinal prestressing. There have been three types of segmental bridgesbuilt in Texas. Span-by-span and balanced cantilever are best differentiated by theirconstruction method and cable-stayed by its use of overhead cables to support the segments.Cable-stayed bridges are discussed in Chapter 9, Section 8 of this manual.

Span-by-Span Construction

Span-by-span construction is most economical in bridges with numerous spans (20 or more)ranging from 80 to 135 ft., with 150 ft. spans possible. A construction cycle is begun byplacing precast segments on erection girders and grading to proper alignment. Epoxy isapplied to match cast joints, and longitudinal post-tensioning is installed to form a load-carrying girder. Generally, a full-depth, cast-in-place closure is cast over the pier andtendons are made continuous between spans; however, a new simplified method has beenused successfully for U.S. 183 (see below). Erection girders are lowered slightly to set thesuperstructure on permanent bearings. See Figure 5-11 for general observations regardingspan-by-span construction.

Balanced Cantilever

Balanced cantilever construction is suited for bridges with main spans between 250 and 800ft. Side spans are typically 55 percent to 70 percent of the main span. This type ofconstruction is characterized by the superstructure erection originating at the pier andprogressing outward in both directions. Cast-in-place or precast segments are placed andpost-tensioned in a symmetrical manner to balance the forces on the pier. Constructioncontinues from all interior piers until the cantilevers connect to form a continuoussuperstructure. See Figure 5-10 for general observations regarding balanced cantileverconstruction.

Projects

♦ In 1973, Texas completed construction of the John F. Kennedy Memorial Causeway inNueces County, the first precast segmental concrete box girder unit in the United States.It was a very modest 100-200-100 ft. unit over the Intracoastal Canal containing twolines of single-cell precast segments erected as balanced cantilevers. Prestressingtendons were located in the concrete and anchored within the web of the box.

♦ In 1975, a 200-290-200 ft. segmental unit was designed as an alternate to a steel plategirder unit, again over the Intracoastal Canal, at High Island, Texas. It was to have been


Spans


erected as cantilevers with false bents in the end spans; however, steel plate girderswere chosen by the contractor.

♦ In 1984, a 320-640-320 ft. balanced cantilever over the Neches River was designed asan alternate to a strutted steel plate girder unit or to a contractor-designed option. Thelow bidder exercised the design option and has constructed a cable-stayed unit usingsegmental concrete box sections.

♦ In 1985, the first segmental contract was awarded in a massive rehabilitation program ofdowntown interstate freeways in San Antonio, Texas. This was big-time segmental,born of an intense desire for aesthetics and minimum interference of construction withexisting traffic. Eventually, there were about 3,000,000 square feet of precast concretebox girder bridge, erected span-by-span. Both internal tendons, where the entire ductand anchors are encased in concrete, and external tendons, where the anchors anddeviators are cast in concrete but most of the duct is outside the concrete in the void ofthe box, are used. The type of segmental design was established by competitive biddingon the first three contracts on separate sets of details prepared by two different privateengineering firms. Three subsequent contracts were for similar designs and detailsprepared by the Bridge Design Section.

♦ An elevated portion of U.S. 183 in Austin, Texas, has recently been completed usingsingle-cell segmental box sections erected span-by-span, except for the connectionsover IH 35, which were erected by balanced cantilever. The span-by-span portion is thefirst known project built entirely with semi-continuous joints over the piers (two testspans in the final San Antonio “Y” project were built this way). A series of simplespans were built, and expansion joints were placed at approximately every third pier.At the two interior piers of a unit, only a thin (8 in.) slab is cast between the endsegments. This semi-continuous joint reduces design and construction complexity withno increase in materials. Flexible neoprene pads allow the entire slab/bearing system toact as a hinge.

♦ An Intracoastal Canal bridge in Brazoria County near Surfside Beach, Texas, wasconstructed incorporating segmental box sections in the design and was erected bybalanced cantilever. While the approach spans utilized conventional AASHTO Type VI(Mod) Beams, the main span, with a length of 350 ft. and a width of 50 ft.-10 in., wasconstructed with cast-in-place segmental box sections. The superstructure depth rangesfrom 8 ft. to 19 ft., with a maximum vertical clearance of 75 ft. to the waterway below.

Design and Construction Issues

Design is complicated. The effects of shrinkage, temperature, creep, and prestress loss mustbe analyzed in coordination with the erection sequence. Sophisticated load history and time-dependent computer programs are a virtual necessity. Geometric design must also beconsidered in terms of the effects on segment casting and erection.

Successful construction of precast segment projects depends, to a great extent, on the skill ofexperienced casting yard personnel. The final geometry of the roadway is set in the castingoperation. Sophisticated geometry programs are essential in the casting operation to accountfor horizontal and vertical curves, camber, deflection, superelevation, and errors from the


Spans


casting of the adjacent segment. The quality of the cast segments will determine the ease offit during erection, and the finish of segments will greatly affect the final appearance of thebridge.

In general, the performance of these bridges has been very good. Problems have beenoccasional and unusually minor. The JFK Causeway Bridge in Corpus Christi, Texas, hasperformed extremely well for nearly 30 years. Some cracking has been repaired in theanchorage zones of two segmental bridges designed with empirical methods. The currentstrut-and-tie method in the standard specifications has recently been used to successfullydesign anchorage zones.

Aesthetics are in the eye of the beholder, but it is generally conceded that the San Antonioelevated structures, with their narrow spine girder and wide overhanging wings, enhance theappearance of the area. Wide boxes with sloping webs that extend to the edge of theroadway are also quite attractive but are more expensive and, therefore, very rare. Single-column piers are often the substructure of choice because of their harmony with the spineand openness beneath the structure.

The Austin structures also have the wide overhang feature and are further highlighted byspecial treatment of the substructure.

Current Status

This type of construction is considered appropriate for structures requiring the most pleasingappearance. Overhead and mobilization expenses make them cost-prohibitive for shortstructures. It should also be considered for spans between 300 and 800 ft. The BridgeDesign Section is capable of design and plan preparation for these bridges.


It is highly recommended that the Bridge Design Section be consulted early in the planningstage.

Geometric and highway design features can often be satisfactorily modified to greatlysimplify construction of segmental concrete bridges. Ramp entrance to or exit from thebridge is undesirable but can be accommodated. There are acceptable adjustments that canbe made to the customary ramp configurations that minimize complications of segmentalconstruction.

Location of bents and span lengths deserve careful consideration. The number of straddleand cantilever bents over lower roadways needs to be minimized for aesthetic reasons.Similar span lengths are desirable for constructibility.

Structural design is controlled by the AASHTO Guide Specification for Design andConstruction of Segmental Concrete Bridges.18 Analysis has been performed using thecomputer programs Bridge Designer II and ADAPT. The Bridge Design Sectionrecommends the following:


Spans


♦ Epoxied joints

♦ A combination of internal and external tendons

♦ Provision for future installation of additional tendons

♦ Two-course asphalt surface treatment and asphaltic concrete overlay on the deck.

Chapter 7 — Superstructure Design Section 16 — Prestressed TxDOT Box Beam Spans


Section 16 Prestressed TxDOT Box Beam Spans

Background

The TxDOT box beams originated in the lower coastal counties in the late 1960s. TheCorpus Christi District had assumed responsibility for constructing a park road along abarrier island, and sought an economical system that would be durable in a salt environment.Pan form spans, the economical favorite, were showing signs of severe deterioration.

A prestressed beam fabricator was marketing a 20 in. deep box beam in the surroundingcounties. Although the first bridge constructed with these box beams was a Goliad Countybridge in the late 1950s, the system was not brought to the attention of the state until the late1960s. The Corpus Christi District decided that box beams would be appropriate for bridgesover the cuts and channels along the island. The Bridge Division prepared the plans and, inthe process, developed the TxDOT 20 in. box beam.

The earlier county box beams had been joined together laterally by welding embedded steelattachments. The Bridge Division chose to use a large cast-in-place concrete shear key withtransverse reinforcing bars threaded through the boxes and bolted for lateral restraint. Thebars were replaced with 1/2 in. seven wire strands since fabrication and constructiontolerances in the hole location prevented a large bar from being threaded consistentlythrough several boxes. The box beams were erected side by side, the transverse strandsthreaded, shear key and bent closure concrete placed, the strands stressed, and the roadwaysurface sealed with a two-course asphalt surface treatment, which was covered by 2 in. ofasphaltic concrete pavement (ACP).

Deeper box beam sections were developed and distributed in a 1975 set of standard details.Box beam depths of 20 in., 28 in., and 34 in. were included in the standards, and 40 in.boxes were designed but not distributed. The large shear key was maintained by using theType A prestressed I-beam shape in the sides of the additional depths.

The transverse post-tensioning strands were initially spaced at 25 ft. increments. The 1975standards also included a detail for the bent closure concrete which essentially locked upeach span. The beams were placed on 2 in. wide by 1 in. thick bearing pads. Shear keyconcrete was then placed between the ends of the beams and under the beams at the interiorbents. The abutment backwall was cast flush with the ends of the beams with the backwallconcrete placed under the beams as well.

These bridges experienced some rather significant problems. Longitudinal cracks usuallyappeared in the ACP deck over each shear key. Other sizeable cracking was observed on afew bridges. A box beam bridge in Fort Worth contained large longitudinal cracks in thebottom flanges of the boxes, as well as large diagonal cracks at the ends of the exteriorbeams. Some transverse tendons corroded and failed under stress.

The Bridge Division revised the 20 in. box beam details and released a new set of 1984standards in an attempt to solve these problems. Additional reinforcing was added to thebox beams. The closure pour at interior bents modified to allow for expansion joints. The



ends of the beams were notched to permit the installation of an armor joint or sealedexpansion joint with a cast-in-place end block. The box beams were placed on twoelastomeric bearing pads at expansion joints.

Despite these changes, problems continued with some bridges. The longitudinal deck crackscontinued to appear. The box beams on some bridges tended to rock transversely on thebearing pads. The Houston District experimented with a reinforced concrete slab on theboxes instead of the ACP overlay. They used slabs from 4 in. to 5 1/2 in. thick, both withand without the shear keys. Houston also switched to a four-pad system, utilizing twoelastomeric pads at each end of the box beam.

The combination of the concrete slab with the shear keys performed well. However, thefour-pad system proved to be somewhat problematic. The beams still had a tendency torock slightly on the pads. This rocking was especially pronounced on heavily skewedbridges. Ultimately, a three-pad system was adopted for box beams.

Box beams have become popular for building bridges where speed of construction orminimum section depth are critical. Span-to-depth ratios of 30 can easily be attained withACP, which is currently the highest possible with precast members. Spans of 118 ft. havebeen constructed with 40 in. box beams. Figure 5-6 shows a typical section through a boxbeam bridge, and includes span capabilities and approximate superstructure depths.

Stage construction can often be expedited by using box beams. The deck can be completedquicker than with beams and cast-in-place slab, allowing first-stage detour traffic to berestored to normal conditions sooner. The cost of box beam spans is considerably more thanfor other prestressed beams, but the advantages sometimes outweigh the additional cost.


Fabrication has been the leading problem with prestressed box beams. The reinforcingdetails were intricate, which promoted conflicts because of bar fabrication tolerances. Theusual potential for honeycomb also existed, along with special problems at skewed box endsdue to camber on release. In addition, forming the void posed a significant fabricationproblem.

The void in each box beam was initially formed with stay-in-place cardboard or styrofoamblocks. The void forms were extremely difficult to hold in position as the concreteplacement proceeded. The void tried to move laterally when the concrete was placed on oneside, and then was forced upward by buoyancy when both sides and the bottom were placed.This often resulted in slab and/or wall thickness that was beyond the specified tolerances.Attempts to circumvent the lateral movement by placing the concrete on each sidesimultaneously often resulted in air voids in the middle of the bottom slab with little or noslab thickness.

TxDOT encouraged sequential concrete placement in 1983 by permitting cold jointsbetween the slab and walls. However, some fabricators continued to place concretemonolithically because of the stringent requirements for cold joint cleanliness. TxDOT later



revised the casting requirements to require a two-stage monolithic pour. The additional costof box beam bridges is primarily in the fabrication of the beams.

Current Status

Standard drawings for box beams are not yet available. Details for 20, 28, 34 and 40 in.deep box beams have been developed and are available from the Bridge Division.Construction details, rail anchorage details, and hold down details are also available. Thesesheets have not been released as statewide standards and therefore must be signed and sealedby a registered professional engineer.

All box beam bridges must have concrete shear keys, regardless of whether a reinforcedconcrete slab or an ACP overlay is used. End diaphragms are required at the end of eachspan. Every beam has a 12 in. wide by 7 in. deep block out at each end to facilitate theinstallation of the diaphragms. The fixed slab joints in a continuous unit have a singlediaphragm cast across both block outs.

Transverse post-tensioning is required with ACP overlay. The strands are located at amaximum spacing of 10 ft. along the length of the span. The tendons are either 1/2 in.Grade 1860 strand or 5/8 in. Grade 1030 threaded bar, and are covered with a seamlessplastic sheath. Interior diaphragms are required in the exterior boxes at the location of thetendons. A 1 in. diameter PVC pipe sleeve is threaded through the tendon holes to protectthe strands from possible corrosion and to allow for possible replacement of damagedtendons in the future.

A three-pad system is currently used with box beams. Typically, the forward station end ofthe beam sits on a single elastomeric bearing pad, while the back station end sits on twosmaller pads.

Box beams are fabricated using a two-stage monolithic casting. The bottom slab is cast inthe first stage, and the sides and top are cast in the second stage while the slab concrete isstill plastic. In addition, cardboard void forms are no longer permitted. All interior voidsmust be formed with polystyrene. Void drain holes are installed at the corners of the bottomslab during fabrication.

Details and section properties for all available standard prestressed TxDOT box beams areshown in Figure 7-16, Figure 7-17, Figure 7-18, and Figure 7-19.


♦ The use of ACP overlay on box beam bridges is discouraged. A cast-in-placereinforced concrete slab is recommended. The slab should have a 5 in. minimumthickness, typically at the center of span (or at center of bearing in situations such as sagvertical curves).

♦ The slab should be Class S concrete, reinforced with #5 bars at 12 in. maximum spacingboth transversely and longitudinally.



♦ Slab overhangs should be avoided. The box beams and gap sizes should be chosen sothat the edge of the slab corresponds to the edge of the top flange of the exterior beams.

♦ The shear key and end diaphragm concrete should be Class S concrete. The enddiaphragms must be cast integrally with the slab. The shear keys may also be castintegrally with the slab.

♦ Bridge skew should be minimized as much as possible. Differential camber and torsionproblems and the tendency of the box beams to rock on the pads become morepronounced as the skew angle increases. Bridge skews exceeding 30 degrees should beavoided.

♦ Box beams are not appropriate for use on curved structures. The complexity of thegeometry required to frame the bridge increases dramatically as the degree of curvatureexceeds 1 degree or 2 degrees. The use of box beams on flared structures should beavoided.

♦ The box beam arrangement should allow a minimum gap of 1/2 in. between boxes and amaximum gap size of 3 in.

♦ 5 ft. boxes should be used as exterior beams when the roadway width requires acombination of both 4 ft. and 5 ft. boxes.

♦ Rail dead load is typically distributed equally among the three exterior boxes when aspan consists of six or more box beams.

♦ Live load should be distributed laterally according to the latest AASHTO Specificationfor concrete beams used in multi-beam bridges.

♦ Concrete strengths for box beams should be limited to 6,500 psi at release and 8,500 psiat 28 days. The required concrete strengths are determined according to the limitationsfor concrete compression and tension stress given in the current AASHTOSpecifications.

♦ Prestressing strands are typically 1/2 in. diameter, 270 ksi low-relaxation strands.

♦ Stresses at the end of the beam are controlled by debonding strands. No more than 75percent of the strands should be debonded. The maximum debonding length should bethe lesser of one-half the span length minus the maximum development length asspecified in the current AASHTO Specifications, 0.2 times the span length, or 15 ft.Strands are typically debonded in 3 ft. increments.

♦ Dowels are no longer used with box beams. Lateral restraint is provided by 12 in. wideby 7 in. tall ear walls located at the ends of each abutment and interior bent cap. A1/2 in. gap is provided between the ear wall and the outside edge of the exterior beam.

♦ Longitudinal restraint at interior bents should be provided only when a continuous unitexceeds four spans in length, or when the beam grade exceeds 6 percent. A detail forthe longitudinally restrained, or “fixed,” interior bents is available from the BridgeDivision. Longitudinal restraint is provided at interior bents only.

♦ Beam hold downs should be used on water crossings when the superstructure could besubjected to pressure flow. The hold downs are typically placed at the center of thejoint between the exterior beam and the first interior beam on both sides of the structure.The hold downs may be moved to the second interior joint for heavily skewed bridges



(approximately 25 to 30 degrees). A minimum gap of 1 1/2 in. is required at a jointwhere a hold down is located.

♦ Bearing seats are not used with box beams. The pads sit directly on the top of the cap.Top of cap elevations should be provided at the points coinciding with the outer edge ofthe exterior boxes at the centerline of bearing. Elevations should also be provided atany intermediate points along the cap, at the centerline of bearing, where either achange in cap slope or change in cap elevation occurs.

Box beams are not vertical, but either parallel the roadway surface when the cross-slopeis constant, or are rotated to the average cross-slope of a span in a transition area. Sincethere are no bearing seat build-ups, the top of the cap must be sloped to match therotation of the beams.

♦ A minimum of three elevation points are necessary for unskewed spans with an evennumber of box beams and a constant housetop slab profile: one at the outside edge ofeach of the exterior beams and a third point at the center of the middle joint. Fourelevations points should be provided for spans with an odd number of beams: one at theoutside edge of each exterior beam, and one at the center of each joint on either side ofthe middle beam.

♦ Framing is more complicated in cross-slope transition areas and skewed bridges. Theorientation of the beams should minimize the variation in slab thickness bothlongitudinally and transversely along the span. This may necessitate stepping the cap atsome joints so that adjacent beams not only have a different slope but also sit at adifferent elevation. Elevation points may be required as frequently as every joint incertain situations. It is possible for the forward half of an interior bent cap to have adifferent elevation than the back half at some locations.

Figure 7-16. 20-inch Prestressed Concrete TxDOT Box Beam Properties (See the followingtable of beam properties. Online users can click here to view this illustration in PDF.)

Beam PropertiesProperty 4B20 5B20

Area (in2) 591.8 717.8Y Top (in) 10.19 10.12Y Bott (in) 9.81 9.88



I (in4) 28,086 35,234Weight (lb/ft) 616 748



Area (in2) 678.8 804.8Y Top (in) 14.38 14.26Y Bott (in) 13.62 13.74I (in4) 68,745 85,370Weight (lb/ft) 707 838





Area (in2) 798.8 924.8Y Top (in) 17.92 17.72Y Bott (in) 16.08 16.28I (in4) 115,655 142,161Weight (lb/ft) 832 963




Beam PropertiesProperty

Beam length 100ft or less4B40 5B40

Area (in2) 918.8 1,044.8Y Top (in) 21.31 21.07Y Bott (in) 18.69 18.93I (in4) 176,607 215,300Weight (lb/ft) 957 1,088

Beam length over 100ftArea (in2) 943.8 1069.8Y Top (in) 21.63 21.36Y Bott (in) 18.37 18.64I (in4) 180,159 219,007Weight (lb/ft) 983 1,114

Chapter 7 — Superstructure DesignSection 17 — Prestressed AASHTO/PCI Box Beam

Spans


Section 17 Prestressed AASHTO/PCI Box Beam Spans

Background

In 1962, AASHTO and the Prestressed Concrete Institute (PCI) published recommendationsfor standard shapes of prestressed concrete I-beams, piling, slabs, and box beams.19 TheBridge Design Section had already developed shapes for I-beams and piling, and did notwant to use slabs and box beams, so the publication had little effect on Texas.

The construction of TxDOT shapes began in 1969, and problems with fabrication andconstruction were soon evident. However, these boxes remained popular for certainconditions.

Departmental research20 was conducted in an effort to develop an economical precastbridge. The research resulted in simplified TxDOT box standards and double tee standards.It also identified the old AASHTO/PCI shapes as probably being simpler to fabricate thanthe TxDOT boxes. Concern about performance of the small shear key delayed the use ofthese shapes with asphaltic concrete overlay. Standard details were prepared, but not issued,for 27, 33, and 39 in. deep AASHTO boxes in widths of 3 and 4 ft., with asphaltic concreteoverlay.

Since 1983, a few bridges have been constructed using special details for the AASHTOboxes. Most of those bridges have been in Houston and all have had reinforced concreteoverlay. Relative costs are indeterminate because of many factors that influence contractorbid prices.

Construction Issues

Construction experience is limited; however, some box problems have been observed. Boxescannot be placed side by side without some increase in the nominal out-to-out width.AASHTO boxes are more susceptible to this lateral growth. Lacking the large shear keyrecess, they have a greater height of side surface to interfere with the adjacent box. Onebridge exhibited unsightly lines on the outside because of the inability of the deck andsubstructure to adapt to this lateral growth. Also, box beam camber can result in the overlaybeing considerably thicker at the span ends than in the center. If the railing rests on the topof the box, it must be increased in height at the span ends to maintain good lines.Adjustments will be necessary also at abutments where finished grade will be above theprescribed roadway grade by the amount of camber minus dead load deflection.

The ultimate problem was cracking in the slab due to lack of adequate shear keys. For thisreason, AASHTO box beams are no longer used.

Current Status

The use of AASHTO box beams is not recommended.

Chapter 7 — Superstructure DesignSection 17 — Prestressed AASHTO/PCI Box Beam

Spans



Design would be similar to TxDOT box beams.

Chapter 7 — Superstructure Design Section 18 — Prestressed Slab Beam Spans


Section 18 Prestressed Slab Beam Spans

Background

In 1962, AASHTO and the Prestressed Concrete Institute published recommendations forstandard shapes for prestressed slabs, along with I-beams and box beams. Round voids wereindicated in the slab sections. None of these were constructed by TxDOT.

In the 1980s the Houston District began experimenting with solid prestressed concrete slabsof 3 and 4 ft. widths, without voids. These slabs could be cast on the same beds as thestandard box beams without the problem of void movement, which is prevalent for theboxes.

Consideration was given to asphalt overlay or concrete slab, overhang or not, shear key ornot, type of bearing and lateral restraint. Usage has grown over the years, especially for off-system bridges.

Current Status

The Houston District went forward with simple rectangular sections without shear keys in 3and 4 ft. widths and with a cast-in-place slab that tied the sections together. The 3 ft. sectionwas later dropped in favor of a 5 ft. wide section. Standards have been developed for thesebeams and are available from the Houston District. The Bridge Division has produced alimited number of custom projects with 4 and 5 ft. wide slabs. Most of these designs werefor an asphalt overlay. They used shear keys and were post-tensioned together. It appearsthat slab beam structures are about the same cost as 20 in. box beam bridges. Dimensions,reinforcing, and strand arrangement are shown for a Houston slab beam in Figure 7-20. Formore information on slab beams with shear keys contact the Bridge Design Section.

Chapter 7 — Superstructure Design Section 18 — Prestressed Slab Beam Spans


Figure 7-20. Houston District’s Prestressed Concrete Slab Beam Dimensions, Reinforcing,and Strand Pattern (Online users can click here to view this illustration in PDF.)

Design Recommendations (Houston’s Slab Beam Only)

No shear key is required. Slabs should be placed on two neoprene bearings at one end andone bearing, centered, at the other end. No longitudinal restraint is required. Lateralrestraint is not required. A 5 in. Class S concrete slab should be used, with one mat ofreinforcing steel, #5 at 6 in. transversely and #4 longitudinal bars placed as shown onstandards.

Design of the beams is according to the current AASHTO Specification, with the followingprovisions:

♦ Slab beams should be Class H concrete of a strength required by the design.

♦ Distribution of wheel loads can be according to the AASHTO Specifications for K=0.7;however, from experience using beam width divided by 5.5 is adequate for mostdesigns.

♦ Computer program PRSTRS1421 will design the slabs as non-standard. Separateproblems are required for each width.

♦ Low-relaxation 0.5 in. strands should be used.

♦ The slab is designed to be composite with the deck. Tie down reinforcing is required.

♦ Shear reinforcing in the slab beam is usually not required, but a nominal amount oftransverse reinforcing is provided on the standard.

Chapter 7 — Superstructure Design Section 19 — Prestressed Single Tee Beam Spans


Section 19 Prestressed Single Tee Beam Spans

Background

Precast single tee shapes are currently available from some fabricators, but this type ofconstruction never became desirable for Texas bridges.

In the early 1960s the Lin Tee, named after its famous originator, was marketed in Texas.One pedestrian underpass was constructed in Waco using this shape. The shape was offeredas an alternate on a reservoir bridge in East Texas, but was beaten by pan form girders,which usually win bidding competition.

A series of standard designs and drawings were prepared in the El Paso District for precastsingle tee bridges. Two twin bridges on IH 10 near Van Horn were constructed in 1968using these details with a reinforced concrete deck. They were also used to replace a bridgelost to flood on a farm-to-market road.

The Bridge Division prepared a set of standard drawings in 1969 for single tees that werenever used. Single tees, composed of Type A beams, with a slab section cast on top by thefabricator, were used for one project in Houston.

That is the total extent of Texas involvement with precast single tees for bridges. Theprimary reason for the unpopularity of single tees was their instability during construction.No maintenance problems have been reported on the few bridges that were constructed.

Current Status

Single tee beams are no longer used and are not recommended.


No design guidelines are needed.

Chapter 7 — Superstructure Design Section 20 — Prestressed Double Tee Beam Spans


Section 20 Prestressed Double Tee Beam Spans

Background

Double tee shapes were suggested by a PCI short span bridge publication in 1975.22 Onefabricator sold double tees for several county bridges around Waco. Texas research in 1982identified these shapes as possibly economical alternatives for short span bridges, and effortsbegan to get them accepted for Texas highway bridges.

The concept was to offer a family of short span bridge types from which the contractorcould select the most economical solution without the administrative complication ofalternate bids. The bid item was each bridge or each square foot of deck, which covered allsuperstructure, substructure, and miscellaneous items. The contractor did not have tocommit to one type until after the letting. This method of bidding is no longer acceptablepractice. Unfortunately for double tees, one member of the family was usually pan formgirders, which are hard to beat for 30 and 40 ft. spans; consequently, there were no doubletee bridges constructed for awhile.

On a long IH 37 river bridge widening project in 1984, double tee beams with reinforcedconcrete overlay were selected over pan form girders and prestressed box beams. The bridgehad slight span length variations and flaring width sections which placed pan form girders ata disadvantage. Also, there was enough volume to justify the cost of forms for the doubletees. Unfortunately, the forms obtained by the fabricator could only be used in the range of30 ft. span length and 22 in. depth.

Shortly thereafter, a similar design was the only alternative on another long river bridgewidening in the same district. This time, another fabricator cast the job in a set of forms thatare more adjustable.

Standard drawings for precast prestressed concrete double tee spans with asphaltic concreteoverlay were completed by the Bridge Design Section in 1985. Efforts continued to fostercompetition for pan form girders. Finally, in 1987, four small projects in the Corpus ChristiDistrict were constructed with double tee beams at the contractor’s option. Meanwhile, afew more small bridges were constructed with double tees and reinforced concrete overlay.

Design Issues

Design problems concern the choice of shape, configuration of beam flange connectionshear key, details of deck joint, and type of overlay.

The PCI shapes were adopted because of a better section for bridge loading, although nofabricator in Texas had forms. Another shape was included in the standards at the request ofa fabricator who owned those forms, but they went out of business before getting a bridgejob. Beam flange connection shear keys are subject to intuition rather than design. Plateswelded to imbedded plates at 10 ft. maximum spacing were used initially. (In 1997, TxDOTreduced this spacing to 5 ft., to reduce field-observed longitudinal deck cracking over flange



joints.) Field performance will check the design enhancements. Deck joints were detailed tocontrol leakage better than the early box beam designs. The first standards had asphalticconcrete overlay only. Oklahoma has reported unsatisfactory performance of double teeswith ACP overlay where de-icing salt is used. The greater volume of Texas double tee usagehas had reinforced concrete overlay, which Oklahoma now uses as standard practice.

Possible span-to-depth ratios are not as great as for box beams and only slightly greater thanpan form girders for 30 and 40 ft. spans. In the 50 to 60 ft. span range, the depth of doubletee spans is about equal to Type A beams with reinforced concrete decks.

Fabrication Issues

Fabrication problems, once the commitment is made to buy forms, have been few. One ofthe few problems concerns the fact that some fabricators have flexible flange bulkheadlocation methods, where others accommodate only the TxDOT 6, 7, and 8 ft. nominalwidths. This caused some fabrication problems when metric width roadways and consequenthard metric converted flange widths were specified.

Construction Issues

Construction problems have not been reported, but should be about the same as for boxbeams in the range of span and skew that have been used to date. The sections are slightlyheavier than corresponding depth box beams, but the double tees are wider, thus requiringfewer pieces per span. There are four bearings under each section, which gives a potentialfor teetering, but the shape is less stiff torsionally and better able to adjust. Shear keyinstallation is a potential problem because of placement tolerances and differential camber.Total camber is theoretically slightly greater than corresponding depth box beams, whichcan cause variation in overlay depth and possible railing alignment problems.

Maintenance Issues

Maintenance problems have now been observed in some bridges constructed by the1980s-era standard details. Cracking in the asphaltic concrete overlay, and, to a lesserdegree, in the concrete decked versions has occurred. The 1997 connector spacing reduction,along with additional improvements based on recently completed research should reduce orvirtually eliminate this cracking. Hopefully, the earlier cracking will not cause severeproblems since de-icing salt will not be used where most of these bridges have beenconstructed. Diagonal stem cracking has been observed in some tee beams fabricated fromthe obsolete standards, in the vicinity of the end diaphragm block-outs. This problem hasbeen corrected in current details by reducing the block-out depth to be no deeper than thebeam flange. The older flange shear connection between beams was found to be difficult toinstall and somewhat ineffective evidenced by the above-mentioned cracking in the ridingsurface. Current details have improved constructibility in this area also.



Current Status

Standard details are currently available for 24 ft. roadways in both an asphaltic concreteversion as well as a Class S (Mod) concrete decked version. (Class S (Mod) reflects areduction in coarse aggregate size to improve consolidation in the flange shear key.)However, it is recommended that the concrete decked version be used to avoid serviceabilityproblems associated with the increased deck cracking in the asphaltic version. Standarddetails for additional roadway widths will be produced and made available on the TxDOTwebsite in the future.

There could conceivably be a situation where double tee beams with an asphalt overlay aredesirable. In this event, be careful to use a two-course surface treatment under the 2 in.overlay. When asphaltic overlay must be used, the designer should consider specifying a 2ft. width of joint reinforcing mesh (“Poly-Guard” or “Pave Prep” products, for example)over all longitudinal joints to reduce deck cracking. This, of course, adds a moderate amountof expense to the project, but it will reduce maintenance costs and improve serviceability.

It is of considerable importance, also, that the field inspection crews are made aware of thecritical nature of the flange connections in insuring deck serviceability. Currently detailedconnection angles should be individually field fabricated for a snug fit, and quality weldingemployed.

If the designer is considering the use of the asphaltic concrete version due mainly to aperceived construction time savings factor (i.e., assuming less concrete curing time for decksurface before traffic can be placed on it), it should be noted that the asphaltic concrete deckoption includes concrete diaphragms at the ends of units that are subject to curing andstrength requirements similar to the concrete deck option. The construction time savings forthe asphaltic deck option compared to the concrete deck option is more in terms of a fewdays rather than weeks under previous TxDOT criteria for strength and curing requirements.

Bearing pads for a range of span lengths are specified on current TxDOT standard sheetsDTBMD-S and DTBMD-O. Pad sizes from previous standards should be avoided to preventproblems associated with undersize bearing surface areas. Field observations of olderstructures indicated numerous instances of pad misplacement, or “hanging out” from understem edges or over cap edges. Pad design criteria for max unit lengths to accommodatethermal expansion/contraction movement is specified on the previously mentioned standardsheets. The maximum beam slope without special beam bearing considerations or end“pinning,” is 5.5 percent. Also detailed on these standards, is a “Type A” deck joint thatprovides an economical alternative to other methods for sealing joints at the ends of units upto 120 ft. in length.

Standard shapes are recommended. Standard shapes and section properties can be found inFigure 7-21.



Figure 7-21. Double Tee Beams (See the following table of beam section properties. Onlineusers can click here to view this illustration in PDF.)

Double Tee Beam Section PropertiesBeam Type Width

(ft)Depth

(in)Yb(in)

Yt(in)

Area(in2)

I(in4)

Weight(plf)

AsphalticConcreteOverlay

6T22 6.00 22.0 14.81 7.21 715 26,572 7457T22 7.00 22.0 15.19 6.81 787 27,937 8208T22 8.00 22.0 15.51 6.49 859 29,110 8956T28 6.00 28.0 18.86 9.14 804 51,577 8377T28 7.00 28.0 19.36 8.64 876 54,286 9128T28 8.00 28.0 19.79 8.21 948 56,617 9876T36 6.00 36.0 24.25 11.75 908 99,647 9467T36 7.00 36.0 24.90 11.10 980 104,965 1,0218T36 8.00 36.0 25.45 10.55 1,052 109,585 1,096

ReinforcedConcrete Slab

6T21 6.00 20.5 13.62 6.88 603 21,140 6287T21 7.00 20.5 14.00 6.50 657 22,292 6848T21 8.00 20.5 14.33 6.17 711 23,283 7406T27 6.00 26.5 17.51 8.99 691 42,511 7207T27 7.00 26.5 17.99 8.51 745 44,881 7768T27 8.00 26.5 18.42. 8.08 799 46,942 8336Y35 6.00 34.5 22.71 11.79 795 84,325 8287T35 7.00 34.5 23.32 11.18 849 89,017 8858T35 8.00 34.5 23.85 10.65 903 93,159 941




Standard double tee shapes are not resident within the PSTRS 14 program, but design caneasily be accomplished using the non-standard option and inputting the section properties aslisted on the DTB-S or DTB-O standard sheets. Designs for standard span lengths from 30to 60 ft. are tabulated on the DTBSD-S or DTBSD-O standard sheets, which are available onthe TxDOT web site. Basic design rules for other prestressed members apply also to doubletees, along with the following considerations:

♦ Reinforced concrete overlay should be 4.5 in. thick (min) with #5 at 6 in. spacing,transversely in the top layer and #4 at 9 in. longitudinally in the bottom layer.

♦ Since the specifications are unclear, a conservative live load lateral distribution factor(derived from recent research and in-house studies) of S/5.25 (for wheel load) has beenadopted for beams with either reinforced concrete overlay or asphaltic concrete.

♦ The distance from outside edge of deck slab to centerline of outside tee stem should notexceed 1.0 ft. for both deck option types, to insure adequate rail impact strength. Thisthen makes it mandatory to place the 6 ft. wide beam on the outside of all bridge spans.It may also be necessary to use this beam at interior locations such as when pedestriansidewalks are specified and the traffic rails occur over interior beams. Possible slabedge treatments are shown in Figure 7-22.

♦ Unless the tee beam spacing can be adjusted, fabricators should be advised to cast thebeams 0.25 in. less than nominal width to allow for lateral growth.

♦ Low-relaxation strands should be used. Fabricators of double tee beams in Texascurrently do not drape strands. Debonding occurs at 3 ft. increments from the beam endssubject to maximum criteria in AASHTO. Debonding is not permitted in the bottomrow of strands, and no more than 50 percent of the strands in one row or in the entiresection may be debonded. Full length debonding to accommodate production lineconsiderations is not permitted.

♦ A total average stem width for both stems must be entered in the “non-standard” sectionproperties area of the program input data to produce proper shear design.

Camber of double tees is significant, but the magnitude of the problem will be limited by therelatively short span capability of the tees.



Figure 7-22. Prestressed Concrete Double Tee Beams (Online users can click here to viewthis illustration in PDF.)

Chapter 7 — Superstructure Design Section 21 — Prestressed Simple I-Beam Spans


Section 21 Prestressed Simple I-Beam Spans

Background

Nothing has been so beneficial to the economy and durability of Texas highway bridges asprecast pretensioned concrete I-beams.

Like many good construction methods, prestressing began in Europe. The first significantprestressed bridge in the United States was the Walnut Lane Bridge in Philadelphia,designed under the supervision of Belgian Professor Gustave Mangel and contracted in1949. William E. Dean, bridge engineer of the Florida State Road Department, was an earlychampion of the method, and the Sunshine Skyway Bridge in Tampa was contracted in1951, using precast post-tensioned beams in the approach spans. Prestressing developmentcontinued in Pennsylvania, Tennessee, and California. Texas’ first attempt at prestressingcame in 1952 when two 30 ft. pan form spans were post-tensioned together to make a 60 ft.prestressed slab and girder span. Cast-in-place post-tensioned bridges soon followed on theDallas expressway.

Highway construction was beginning to escalate, heading toward the interstate boom thatbegan in 1956. Steel beams were showing signs of unpredictable availability and escalatingprices. The arrival of prestressing was very timely.

The Bridge Design Section chose to develop the capabilities of precast prestressed I-beams.The first significant beam bridge in Texas was over Corpus Christi Harbor. It was contractedin May 1956 and contained 2,000 ft. of 40 and 60 ft. prestressed concrete I-beams. Thebeams were of special shape, precast on the job, and post-tensioned. Simultaneously,standard shapes were being developed that were suitable for pretensioning or post-tensioning. The precast concrete industry was eager to help in the development. For severalyears, the standard details were fine-tuned to fit fabrication capabilities and a variety ofgeometric configurations. The basic shapes of these early standard beams have remained thesame through the years.

In the early years of interstate highway construction, prestressed beam spans were offered tothe contractors as alternates to continuous steel I-beam units. It soon became obvious thatsteel could not compete economically. Prestressed beams became the best choice for manycrossover structures and stream crossings. Spans over 65 ft. were avoided for awhile, but thesize and strength of prestressing strands increased and expertise in depressing strandsdeveloped, confidence in longer spans increased. Arrival of the AASHTO Type IV Beam,along with the demonstrated ability of the prestressing industry to produce high-strengthconcrete, allowed spans of 130 ft. to be constructed economically with prestressed beams.

Figure 7-23 gives a history of quantities and bid prices for prestressed I-beams since 1963.Of the bridges for which bids were taken during that period, approximately half containedprestressed concrete I-beams.



Figure 7-23. Bid Letting Statistics (Online users can click here to view this illustration inPDF.)

Standard shapes were established in 1956. Types A, B, and C beams had larger bottomflanges to accommodate more strands. Type A became the AASHTO Type I, but the othertwo have remained proprietary to Texas. Types 36, 48, 54, 60, 66, and 72 were developedsimultaneously. Only Types 54 and 72 have been used to any extent, and only in Texas.Details for AASHTO Type IV beams were prepared in 1968, although there had been someprevious usage in Texas. The AASHTO Type VI (Mod) was first offered on a Texas projectin 1987.



Standard Span and Bent Details

Standard span and bent details were first developed for Types A, B, and C beams in 1957.The beams each had one straight pattern of 3/8 in., 250K strands, and span lengths werelimited to 57.5 ft. Span lengths were tailored to the geometric requirements for interstatehighway crossovers at the time. A new series was distributed in 1960 that allowed onlyType C beams and had spans to 65 ft. with more roadway widths. The beams now had twostraight patterns each and an optional depressed pattern. In 1965 the Type C beam designwas shown as a depressed pattern using 7/16 in., 250K strand with optional designparameters. The fabricator could design his own pattern in 1/2 in., 270K maximum strands.Spans through 80 ft. were covered in five different roadway widths. Sometime between1965 and 1968, end blocks were removed from pretensioned beams. The 1971 standardspans and bents covered Type C and Type 54 beams with spans to 110 ft., five roadwaywidths, and four skews. Standard spans and bents were not provided for AASHTO Type IVbeams, until they were issued in metric units in 1996. An updated English unit standardfollowed in 1999.

Many bridges were constructed to standard span and bent details, but many more requiredspecially prepared details because of non-standard geometry. Currently, standards coverTypes A, B, C, and Type IV beams in five roadway widths and three skews. Beam shapesare the same as those established in 1956 and 1968.

Modified Shapes

Shapes have been modified occasionally for some special purpose. The Type C plus 6 in. ofdepth was used on a few railroad underpasses. Type 54M, the fat 54, was used for a briefperiod for increased span length while matching the Type 54 for perceived aesthetics of thetotal bridge. Economy was usually maintained by allowing standard side forms and astandard soffit.

One notable modified shape project, the Buena Vista and Commerce Street overpass in SanAntonio, was let in 1957 using Type C side forms. Parallel 1,600 ft. bridges carried citystreets over a series of railroad tracks. Span-to-depth ratio needed to be maximized becauseof congested conditions. Type C depth was maintained throughout the bridge, including 90and 100 ft. spans over the main tracks. These beams were spread to a 10 in. web, fabricatedof lightweight concrete, and post-tensioned. A lightweight concrete deck was alsoconstructed. Until recently the bridges were still functioning but the decks are badlydeteriorated. While most of the beams are still sound, the decks have now been replaced.

Post-tensioning was allowed for the standard beams for several years but was seldom used.The contractor for one end of a long ship-channel bridge in Houston cast Type 54 andType 72 beams on the job and post-tensioned them. The other end of the bridge waspretensioned.

Although theoretically unstable, the Type 72 beam had been used for 140 ft. spans. It maybe economical above 100 ft. if depth is unimportant. Consideration of this beam wasrecommended for stream crossing structures with spans between 100 ft. and 120 ft. The



beam was also adaptable to railroad underpasses in the 70 ft. span range. Because of thelateral instability the length was limited to 100 ft.

One experimental bridge in Sherman, Texas, contained two 150 ft. spans that were Type 72beams match cast in three sections and post-tensioned together on the ground beneath thestructure. There were problems with fit of abutting pieces and with gluing the joint, but thestructure was completed. Calculated camber due to prestressing was about twice the amountthat actually occurred. Slab weight deflection left the beams with a slight sag. Thecomplication of this method is considered to outweigh any economical advantage.

Larger span-to-depth ratios can be obtained with cast-in-place concrete units and continuoussteel beams, but economy and ease of plan preparation made prestressed beams the favorite.

Design Issues

Design problems have been few. The key item is stress in the bottom flange at mid-span.The specification allowable has gone from zero to about 400 psi tension and in 1986 wasthreatening to go back again. Loss of prestress affects the bottom stress considerably. Afteryears of design using 20 percent loss and a lot of controversy, the specification wasequipped with a complicated loss formula. Shear considerations also became complicated inthe 1981 interim specification. Concerns about strand fatigue, debonding strands, and shearthreaten to complicate things more. In the design stage, it has not been possible to predictcamber with any consistent accuracy.

Fabrication Issues

Fabrication problems have been many, but they were usually solved through theperseverance of the fabricator, the Materials and Tests Section, and the Bridge DesignSection. Congestion of reinforcing steel in end regions was a problem until equitable detailsevolved from trial and error. Bent bars are kept as small as possible with strict tolerances onbending radii. Some fabricators bend their own reinforcing to maintain control of thetolerances. There have been few problems with prestressing strand.

The condition of the anchorage chucks appears to be the key to eliminating strand breakage,which can be lethal during fabrication because of the extreme length of stressed strand.Some difficulty has been experienced verifying the properties of low relaxation strand.Texas was reluctant to accept certification, so a 30 minute relaxation test was devised to runon lot samples in the Materials and Tests Section laboratory. There has been somecontroversy regarding the accuracy of the test, but it appears adequate to distinguishbetween low relaxation and stress relieved strand.

Proper vibration of concrete is necessary to prevent honeycomb, and external vibrators aloneare not sufficient. Depressing of strands into a trapezoidal pattern was developed throughconsiderable effort and ingenuity by the fabricators. Hold-down hardware must be releasedprior to release of strand tension to prevent damage as the beams move along the soffit.Release of the hold-downs causes the beam to deflect upward. If the number of depressedstrands and the angle of depression is great, the stresses associated with this deflection will



cause cracking in the top of the beam over the hold-downs. This problem has been solved byadding dead weight to the beams before release of hold-downs.

Another problem occasionally associated with upward deflection is corner cracking at thebottom ends of beams as they drag along the soffit during release. However, this is difficultto predict. Dead weights help, and thin slick sheets inserted between beam and soffit beforerelease have been used effectively to prevent further cracking. Cracking in the transfer zoneat the beam ends has always been a possibility. When end blocks were eliminated, particularattention was directed to this problem. Cracking without end blocks was seen to be verysimilar to transfer zone cracking in beams with end blocks. The cracks are usually small andhorizontal at the flange-to-web juncture or diagonal in the web such that they tend to closeunder installation conditions. A system of selective strand wrapping near the ends wasdevised that effectively eliminates this problem. Some horizontal end cracking has beenobserved in the face of bottom flanges. It appears to be associated with a lack of reinforcingsteel in that face and aggravated by prolonged storage on supports located some distancefrom the end of beam. Strand wrapping itself can cause splitting if not properly done. Tapingalone is insufficient. Concrete strength is not generally a problem, although some areas ofthe state have some difficulty. Periodically, even the good areas appear to fall into a slumpin strength. Usually the lapse can be attributed to aggregate strength, cement condition, orweather, but there have been times when the reason remained unknown.

A system of structural review is in place that allows the fabricator some recourse on beamsthat are not fabricated in conformance to the plans and specifications. If the Materials andTests Section inspectors reject a beam for non-conformance, the fabricator may request astructural review. Unless the beam is obviously beyond hope, the deficiencies will berecorded and forwarded to the Bridge Construction Section for review. Typical deficienciesare honeycomb, cracking, misplaced reinforcing steel, and insufficient concrete strength.The Bridge Construction Section, or a design consultant, will make a determination ofadequacy based on comparison of the deficiency with detailed design requirements. Thebeams may be accepted as is, considered acceptable after proper repair, or rejected.Acceptance without repair usually requires a return of a little money to the state asrecompense by the fabricator to justify acceptance of a beam in non-compliance with thespecification.

Construction Issues

Construction problems have occurred mostly with the Type 54 beam. Originally designedfor spans up to 110 ft., the beam was later limited to 95 ft. by TxDOT policy. It had beenobserved that the beam becomes unstable at lengths greater than 102 ft. The beam is 54 in.deep with 16 in. wide top and bottom flanges, which tends to lateral instability in longlengths. These beams are more susceptible to breakage during hauling and on severaloccasions have broken during the erection process, creating a dangerous situation. For sometime, adjustable external stiffening devices (hog rods) have been required for Type 54 beamsover 96 ft. long. This beam also tends to excessive lateral deflection (sweep) after erection.Beams outside of the specification tolerance for sweep must be field straightened and helduntil the diaphragms and/or slab are placed.



Diaphragms have always been a source of irritation to the contractors because of theexpense of forming. Bolted steel interior diaphragms have been allowed for some time and,lately, interior diaphragms have often been omitted. A research project conducted by theCenter for Highway Research (Report #158-1F) concluded interior diaphragms were noteffective.23 Girder impact damage is more severe with the use of interior diaphragms. Inthe late 1980s TxDOT discontinued the use of interior diaphragms except on unusually longand unstable beams, in which case steel interior diaphragms are typically specified. Enddiaphragms are now of such configuration that they may be formed and placed with the deckslab. Thickened slab ends became standard in 1996.

Torsion in the outside beam due to deck slab placement in the overhang was a problem untila system of bracing was devised and required by the specifications. Extreme camber andcamber differential between adjacent beams has caused construction difficulties.

Maintenance Issues

Maintenance problems through 1990 have been few. Leakage of salt-laden water through thedeck joints has caused deterioration of caps and columns, but the high-strength prestressedconcrete appears to be very durable. Deck slabs have deteriorated and been replaced overbeams that are still in good condition. The 1957 standards had a detail that was unsightly.The precast beam stopped about 6 in. short of the deck joint centerline and was completedby a concrete patch cast with the deck. The patch was supposed to be insulated from the topof the cap but, unless careful attention was given, concrete would be cast directly on the cap.This became the bearing under load, resulting in the patch cracking away from the beamend. Also the patch, being deck concrete, deteriorated due to salt exposure. Slightdeterioration of the beams, where the strands are thus exposed, has been noted. A 25-year-old pretensioned county bridge close to salt water near the gulf coast was found to beapproaching the threshold level of chloride 2 in. beneath the beam surface. Embedded largemild steel anchor bars were rusting, but there was no evidence of strand rusting. Speculationwas that strand rusting would be just a matter of time, but proven relief measures have yet tobe found. Lately, severe damage to beams has been noted in a long salt-water crossing atGalveston Bay. This, and other observations, seem to indicate that even prestressed I-beamsare not immune to destruction by salt water.

Highway crossover beams get hit by overheight loads frequently. Depending on how manystrands are broken, the beam may be patched with gunite or dry packed mortar or removedand replaced. Occasionally, a flammable load will burn beneath a bridge. Damage is usuallyconfined to concrete cover over the bars or strands. After one such accident, involving awhiskey truck, two prestressed beams were replaced because of soft concrete in the webnear the beam ends.

Long-Term Performance Issues

There has been no evidence observed in the field to indicate design deficiencies in flexure orshear. There is a bridge known to have existed for 15 years on an interstate highway withprestressed beams that have an inventory rating of H6 due to a recently discovered designerror. No cracking or unusual deflection under load has been observed. Laboratory tests



continue to identify weaknesses in flexural fatigue, shear, and shear fatigue, but they haveyet to show up in the field. What does stand out as a design deficiency is an inability topredict deflections.24 Beam camber, dead load deflection, and longtime compositedeflection are far from consistent with analysis methods currently employed.

Concern over strand fatigue erupted in 1978 when a PCA testing project for the LouisianaDOT produced premature fatigue failures in full-size pretensioned beams. Subsequentextensive research at the University of Texas25 verified that fatigue failure is possible in thestrands of a cracked pretensioned beam subject to repeated flexure. The research alsodemonstrated that an uncracked beam would soon crack under repeated flexure to a stressnear the modulus of rupture of the concrete. Field performance of prestressed beam spanshas given no indication of this problem. AASHTO Specifications have not been revised as aresult of the research.

Current Status

Prestressed concrete beams continue to be used in about 45 percent of Texas’ bridges. Theyclearly result in the most economical and durable bridge within their span capability. Theyare suitable for most geometric conditions and structure types, including widenings andrailroad underpasses. Quality beams are readily available from several competitive sources.Transportation is not a deciding factor in urban areas up to 130 ft. beam lengths. Beams upto 150 ft. have been successfully transported, but at a premium cost. In rural areas, longType IV beams may cause transportation problems.

♦ Type 54 beams are no longer recommended for use. The economy of the Type C andType IV have made the Type 54 obsolete. This beam should only be used to widenexisting bridges that used Type 54 beams.

♦ Type 72 beams are no longer recommended for use. This beam should only be used towiden existing bridges that used Type 72 beams.

♦ AASHTO Type VI (Mod) beams are now economically available in Texas.

♦ Low-relaxation strand is readily available and economical. The standard strand forprestressed beams is 0.5 in. 270K low-relaxation. Low relaxation strand of 0.6 in.diameter is approved and may be accepted in optional designs. It should only berequired for special projects.

♦ Overlength permits are readily available for beams 150 ft. long.

♦ Concrete strengths of 6,500 psi at release of prestress and 8,500 psi at 28 days arefeasible for usual designs. Greater strength requirements should be restricted to specialprojects.

♦ Competition is such that predictions of which fabricator will sell beams for a particularproject are unreliable. Transportation costs are apparently not significant.

TxDOT, through research, has been developing ways to produce higher strength concrete forseveral years. It now appears that prestressed concrete manufacturers can achieve strengthsin excess of 12,000 psi in workable batches of concrete. To utilize this strength effectively,additional prestressing is desirable. To achieve this, Texas and other states have sponsored



extensive research toward the use of 0.6 in. diameter prestressing strands. The result hasbeen approval by the FHWA to use 0.6 in. diameter strands in a 2 in. grid spacing. A recentproject in San Angelo utilized high strength concrete with a concrete strength of 14,000 psito construct a 153 ft. span with Type IV beams. Specific design conditions and experiencewill decide the economy or desirability of using this capability.

Details and section properties for all available standard prestressed I-beams are shown inFigure 7-24, Figure 7-25, and Figure 7-26. Prestressing strand constraints are shown inFigure 7-27and Figure 7-28.

Figure 7-24. Standard Prestressed Concrete Beams – Types A, B, C, and IV (See thefollowing table of beam dimensions and section properties. Online users can click here toview this illustration in PDF.)

Beam Dimensions and Section PropertiesBeamType

Ain

Bin

Cin

Din

Ein

Fin

Gin

Hin

Win

Ytin

Ybin

Areain2

Iin4

Wt/Lflbs

A 12 16 5 28 5 11 3 4 6 15.39 12.61 275.4 22,658 287B 12 18 6 34 5 ¾ 14 2 ¾ 5 ½ 6 ½ 19.07 14.93 360.3 43,177 375C 14 22 7 40 7 ½ 16 3 ½ 6 7 22.91 17.09 494.9 82,602 516IV 20 26 8 54 9 23 6 8 8 29.25 24.75 788.4 260,403 821



Figure 7-25. Standard Prestressed Concrete Beams – Types 54 and 72 (See the followingtable of beam dimensions and section properties. Online users can click here to view thisillustration in PDF

Beam Dimensions and Section PropertiesBeamType

Ain

Bin

Cin

Din

Ein

Fin

Gin

Hin

Win

Ytin

Ybin

Areain2

Iin4

Wt/Lflbs

54 16 16 8 54 5 30 5 4 6 28.47 52.53 493.4 164,022 51472 22 22 11 72 7 ½ 40 ½ 7 ½ 5 ½ 7 38.27 33.73 863.4 532,060 899



Figure 7-26. Standard Prestressed Concrete Beams – Type VI (Mod) (See the followingtable of beam dimensions and section properties. Online users can click here to view thisillustration in PDF.)

Beam Dimensions and Section Properties for Type VI (MOD) BeamsAin

Bin

Cin

Din

Ein

Fin

Gin

Hin

Jin

Kin

Win

Ytin

Ybin

Areain2

Iin4

Wt/Lflbs

40 26 8 72 10 42 3 5 4 13 6 35.54 36.46 940.4 670,351 980



Figure 7-27. Standard Strand Information – Types A, B, C, 54, and 72 (Online users canclick here to view this illustration in PDF.)



Figure 7-28. Standard Strand Information – Types IV and VI (Mod) (Online users can clickhere to view this illustration in PDF.)


The various types of standard prestressed concrete beams and their recommended usage is asfollows:

♦ Type A - Depth 28 in.Used primarily for widening old concrete and steel spans for compatibility with existingdepth. Reasonable span limit is about 50 ft.

♦ Type B - Depth 34 in.Used for widening and for new structures where depth is important. Reasonable spanlimit is 65 ft.



♦ Type C - Depth 40 in.The dominant beam of the group. Over 6 million linear feet used since 1963. Suitablefor widening, grade separations, stream crossings, pedestrian underpasses, and short-span railroad underpasses. Economical span limit about 75 ft. Maximum recommendedspan limit 90 ft.

♦ Types 36, 42, 48, 60, and 66Forms are not available for these beams. They should not be used.

♦ Type 54 - Depth 54 in.Historically the second most popular beam, over 3 million linear feet have been usedsince 1963. However, because of lateral instability, construction problems, and cost,this beam is now considered obsolete.Note: Preferably, this beam should not be used at all. However, it may be used forwidening at existing Type 54 spans. The Type IV beam can also be used to widenexisting Type 54 bridges and may prove more desirable.

♦ Type IV - Depth 54 in.Since 1986 this has been the dominant beam. Over one million linear feet were used1986 through 1988. This is a tough stable beam, and it is recommended for span lengthsup to 130 ft.

♦ Type 72 - Depth 72 in.Because of lateral instability, construction problems, and cost, this beam is nowconsidered obsolete.Note: Preferably, this beam should not be used at all. However, it may be used forwidening at existing Type 72 spans. The Type VI (Mod) beam can also be used towiden existing Type 72 bridges and may prove more desirable.

♦ Type VI (Mod) - Depth 72 in.This is a newcomer. The wide top flange improves lateral stability. The Type VI bottomflange allows a maximum number of strands. The beam has 175 ft. span capabilities buthas been limited to 150 ft. due to handling constraints. This beam is recommended forspans greater than 130 ft.

The suitability of prestressed concrete I-beams for railroad underpasses is severely limitedby the requirement of 18 in. clearance between flanges imposed by one railroad companyand a 12 in. requirement by others. Designers should check with the Bridge Design Sectionfor the latest information.

For grade separation structures the same beam depth should be used for the full length ofstructure, for aesthetic reasons. Stream crossing structures may have different types andsizes of beams if economy so dictates. Beam spacing should be optimized in each span.There is no significant advantage to maintaining a constant beam spacing for the full lengthof structure. Selection of the proper type of beam for a span is a matter of economics.Relative costs should be calculated using current average bid prices for beams and slab.

Concrete strength maximums may be taken as 6,500 psi release and 8,500 psi design, but forlarge volume structures, verification with a prospective fabricator is advisable. High strengthis not achieved without some extra cost, which will be unknown to the designer.Unfortunately for the fabricator, these costs are often lost in the bidding competition.



Beams should be designed for 0.5 in. low-relaxation strands using the computer programPSTRS14. Optional design parameters for maximum top flange stress, bottom flange stress,and ultimate moment due to all design loads should be shown on the plans for each design.The fabricator will retain the option to use other strand arrangements, including straightstrand patterns, stress relieved strand, or 0.6 in. diameter strand, provided the designparameters are satisfied by the prestress and concrete strength selected. Although PSTRS14allows considerable liberty in selecting design controls, the following practice isrecommended for standard designs:

♦ Strands should be added and depressed in the order shown on standard drawing IBNS.See Figure 7-27 and Figure 7-28.

♦ Hold-down points are shown on the standard to be 5 ft. or .05 span length (if greater)either side of mid-span. Fabricators are allowed -0, +2 ft. tolerance from this.

♦ Strand stress after seating of chucks will be 0.75 f 's for low-relaxation strands.

♦ Section properties given on the standard drawing and built-in to the program will beused.

♦ Section properties of the beam should not be increased to account for the transformedarea of strands or mild steel.

♦ Composite section properties should be calculated assuming the beam and slab to havethe same modulus of elasticity (for beams with f 'c < 7,500 psi) with no haunch betweentop of beam and bottom of slab.

Note: The previous two items tend to compensate each other and simplify the designprocess.

♦ Live load distribution should be S/5.5 wheels for moment and for shear.

♦ The program will iterate to the required number of strands using loss calculations inaccordance with the 1989 AASHTO Specifications.

♦ The primary control will be final stress in the bottom of beam at midspan. Tension in

the amount of 6 f c' will be allowed.

♦ The required f 'c is calculated based on the following:a. The compressive stresses under all load combinations, except as stated in (b) and

(c), shall not exceed 0.60 f 'c.b. The compressive stresses due to effective prestress plus permanent (dead) loads

shall not exceed 0.40 f 'c.c. The compressive stresses due to live loads plus one-half of the sum of compressive

stresses due to prestress and permanent (dead) loads shall not exceed 0.40 f 'c.

Final stress at the bottom of the beam at the ends will not be checked.

♦ Release strength will usually be controlled by the compressive stress after release at thebottom center of beam. The effective strand stress after release will be 0.75 f 's-ES-0.50CRs for low-relaxation strands.



♦ The end position of depressed strands will be as low as possible to prevent control ofthe release strength, or higher if necessary, to limit top tension to cf5.7 ′ . Occasionally,release strength will be controlled by end conditions when the depressed strands havebeen raised to their highest possible position.

♦ Required stirrup spacing is calculated for #3 Grade 60 bars according to the AASHTO1989 Specification. Stirrup spacing according to ACI 318-89 is also shown butAASHTO governs. Spacings are given at the bearing, one-half the composite depth andthe twentieth point of the span, where principal tension is likely to control the shearcarried by the concrete. At the tenth points of span, shear at inclined cracking will likelybe the concrete contribution, but usually the nominal maximum spacing will control.The required stirrup spacing at the bearing will always be satisfied by the splittingreinforcement in the transfer zone. The requirement at h/2 should be satisfied from theend of splitting reinforcement to the 1/20 point and so on.This method of shear consideration was adapted from equations introduced in ACI-318-71. It may be complicated, but it is the official AASHTO method and, according torecent research, predicts shear capacity as well as other known methods.

♦ Horizontal shear between beam and slab may be ignored for standard beam details inhighway bridges that are considered to be in compliance with the AASHTO horizontalshear specifications.

♦ Camber is calculated using an adaptation of the computer solution.26 This method hasbeen observed to predict average camber within tolerable limits.More accurate methods may be justified for unusual conditions if the more importantparameters affecting long-term camber can be controlled.27

♦ Deflections due to slab weight and composite dead load are based on the input value ofmodulus of elasticity of the beam (5,000,000 psi is recommended). This should beshown on the plans although field experience indicates actual deflections are generallyless than predicted. Use this deflection times 0.8 for calculating haunch depth.

♦ When precast concrete deck panels are allowed, the beam should be designed using thebasic slab thickness, except in rare cases.This is considered justified by successful testing.28

♦ When stay-in-place metal forms are allowed, the design slab thickness may be used forbeam design. Additional dead load of concrete required because of the corrugations isnot considered.

♦ Thickened slab ends should be detailed at the ends of each simple span or at the end ofeach unit. End diaphragms will usually be an option of the contractor. See Figure 7.29.

♦ Intermediate diaphragms are not required except for erection stability of Type VI (Mod)beams or other beam sizes stretched beyond their normal span limits.



Figure 7-29. Slab End Details – Prestressed Concrete Beam Spans (See followingexplanatory notes. Online users can click here to view this illustration in PDF.)

Explanatory Notes for Figure 7-29Many structures have been built successfully without a thickened slab end at inverted Tbents. The stem of the inverted T is assumed to act as support for the slab edge. However,this remains a controversial issue. On several projects, particularly those with wide invertedT stems, a modified thickened slab end has been used. Its length is usually half of the stemplus 4 feet into the span. This tends to be a local preference issue and the designer isencouraged to contact the Bridge Design Section or the district bridge design office forrecommendations.

Chapter 7 — Superstructure DesignSection 22 — Prestressed Cantilever/Drop-In I-

Beam Spans


Section 22 Prestressed Cantilever/Drop-In I-Beam Spans

Background

To increase the span length capability of prestressed concrete beams, a few bridges havebeen designed with sections extending past the interior bents to support a simple span. Thusthe span capability of a particular beam type could be increased by the sum of the twocantilever lengths and the structure remain determinate.

The first design used variable depth pretensioned girders as the cantilever member. Theywere to be cast and hauled upside down, righted during erection, and connected across thecenter by standard prestressed beams notched to fit the cantilever connection. This produceda 110 ft. span with Type C beams. It was offered as an alternate to a continuous steel girderunit on a Brazos River bridge, but was not selected for construction.

The first unit constructed had cast-in-place reinforced concrete cantilevers with drop-in 75ft. Type C beam spans. This project resulted in an 87.5-100-87.5 ft. unit across the BosqueRiver at Iredell.

The next variation had 85 ft. cast-in-place concrete girders cantilevered past the interiorbents with 90 ft. Type C beam drop-in. This was a 62-116-62 ft. interchange structure inWaco.

An interesting and aesthetically pleasing variation was constructed on a long interstateelevated highway in Temple. Notched Type C beams were used, but the matching cantileversections were cast-in-place bent caps, flush with the bottoms of beams. Every third bent wasa drop-in situation with deck expansion joints. The two interior bents were cast around thebeam ends while they were supported by falsework.

The type that followed, and was used in several bridges around the state, consisted ofstandard shape pretensioned beams cantilevered across the interior bent and notched tosupport the same beam shape as a drop-in. The usual beam was Type IV and the longestspan achieved was 165 ft. with a 135 ft. drop-in span. Later, cantilever/drop-in Type 72beams were chosen by the contractor over 150 ft. simple spans using Type 72 or Type VI(Mod) beams, for a coastal pleasure boat channel crossing near Kemah.

The most severe use of this type of bridge was the Cypresswood Drive Overpass, designedby others, in Houston. The spans were 113-188-123 ft. using Type VI (Mod) beams with a140 ft. drop-in span. The system has also been used for widening continuous steel girders.

Design Issues

Design of these units is tedious since the cantilever section pretensioning must be locatedlow to resist positive moments within the span and high to resist negative moment at theinterior bent. The problem is compounded by the fabricator who, in order to check the work,


Beam Spans


invariably wants to redesign the strand pattern, creating another task for the designer. Thenotched end has always been a concern of designers, since the detail is not subject toaccurate rigorous analysis. A conservative detail was suggested by the Fort Worth District,successfully laboratory tested at the University of Texas,29 and became the official detail forthe state. A sample of this detail is shown in Figure 7-30, Figure 7-31, and Figure 7-32.Early designs had deck expansion joints at the cantilever ends, but later units were madecontinuous, in the deck slab only, by continuing the longitudinal slab reinforcing through thejoint. Longitudinal reinforcing over the interior bents was designed for negative live loadmoment.

Figure 7-30. Example of Cantilevered I-Beam Details – Type IV Beam (Online users canclick here to view this illustration in PDF.)


Beam Spans


Figure 7-31. Example of Drop-In I-Beam Details – Type IV Beam (Online users can clickhere to view this illustration in PDF.)


Beam Spans


Figure 7-32. Example of Cantilever/Drop-In I-Beam Details – Bearing Seat and StrutAssembly (Online users can click here to view this illustration in PDF.)

Fabrication Issues

Fabrication was complicated by the need to deflect a group of strands downward in the endspan and upward over the cantilever support. The required number of strands was usuallylarge and concrete strengths on the high side. Embedment of reinforcing and steel platesused at the notched ends created difficult forming and concreting problems.


Beam Spans


Construction Issues

Construction posed no particular problems except for one unit that developed pronounceddeformation due to prestressing in the cantilever section. Elastomeric bearings requiredadjustment to fit the erected slopes of distorted cantilever and drop-in bearing surfaces. Slabforming adjustment was also aggravated. Problems continue with misalignment of the notchsurfaces, but this can be mitigated with proper forethought.

Prestressed concrete deck panels are usually not allowed in negative moment regionsbecause of insufficient clearance for large-diameter longitudinal slab bars over the cantileversupport.

Current Status

Cantilever/drop-in prestressed concrete beam units are no longer recommended. The use ofHPC simple spans, Type VI (Mod), or structural steel spans should be more economical andare preferred from a long-term durability and maintenance standpoint. However, thefollowing information is provided if the need for cantilever /drop-in prestressed concretebeam units is unavoidable.

Cantilever/drop-in prestressed concrete beam units using Type IV beams may be used forspans between 130 ft. and 150 ft. and Type VI (Mod) beams may be used for spans between150 and 190 ft. where economy and avoidance of structural steel is more important thanaesthetics.

Various span arrangements that have been used are shown in Figure 7-33.


Beam Spans


Figure 7-33. Cantilever/Drop-In Precast Pretensioned Concrete Beam Spans (Online userscan click here to view this illustration in PDF.)


If there is an appropriate occasion to use this structure type, the design should closely followthe AASHTO Specifications with the following considerations:

♦ Type IV beams or Type VI (Mod) beams should be used for grade separation structures.

♦ End blocks will be required on the notched ends of prestressed beams to provide spacefor bearings and notch reinforcement.


Beam Spans


♦ Notch reinforcement should be similar to the detail shown in Figure 7-30, Figure 7-31,and Figure 7-32 and should conform to the requirements in “Optimum Design orReinforcement for Notched Ends of Prestressed Concrete Girders.”30

♦ Standard elastomeric bearings may be used at the drop-in supports. End bearings mustbe designed to accommodate the movement caused by temperature change. Interiorbearings should be designed for the heavier-than-standard loads that will exist.

♦ Deck expansion joints are not recommended at the notched ends. Instead, the potentialjoint should be restrained by continuous longitudinal deck slab reinforcement. A deckconstruction joint and overhang chamfer, similar to details shown in TxDOT StandardIBMS, should be required to control the crack.

♦ Deck slab reinforcement over the cantilever support should be sized to help resistultimate negative moment. Grade 60 steel should be used.

♦ Concrete strength for the cantilever beam will usually be controlled by compression inthe bottom flange over the support. Design concrete strength should generally belimited to 8,500 psi.

♦ Stress at tenth points of the span must be calculated to ensure adequate distribution ofprestress. Cantilever beams should be considered pinned at the notched ends.

♦ Computer program BMCOL5131 can be used to calculate moments, shears, anddeflection due to external loads in the cantilever section.

♦ Moments, shears, and deflections due to the eccentricity of prestressing can also beapproximated with BMCOL51 by inputting end moments and effective vertical forcesat strand deflection points.

♦ Prestress losses may be hand calculated separately at each maximum moment location.

♦ Live load distribution should be S/5.5 wheels to each beam.

♦ Shear effects should be investigated according to the AASHTO Specification.

♦ Camber in the cantilever section may be approximated using BMCOL51 and themultipliers in “A Rational Method for Estimating Camber and Deflection of PrecastPrestresseed Members,”32 and “Time Dependent Deflections of Pretensioned Beams.”33

Chapter 7 — Superstructure DesignSection 23 — Prestressed Continuous for Live Load

I-Beam Spans


Section 23 Prestressed Continuous for Live Load I-Beam Spans

Background

Most problems with simple prestressed concrete beam spans are caused by leakage throughthe deck joints at the ends of each span. In an attempt to circumvent this problem,continuous for live load designs were introduced in the early 1960s.

Continuity was provided by mild steel reinforcing in the deck slab across the bent with thebeams connected in the compressive zone through cast-in-place diaphragms. To resistpullout of the beam bottom due to shrinkage, creep, and temperature-induced positivemoment, large reinforcing bars from each adjacent beam were extended past each other andbent upward in the diaphragm. This detail was considered necessary because of the limiteddistance between adjacent beam ends. Later designs had an inverted tee bent with the stemcast around the beam ends. With the stem width available, prestressing strands from eachadjacent beam were sometimes extended for anchorage. Some units with inverted tee capshad the flange of the cap reinforced to carry beam weight. Others required the flange to besupported on falsework until the beams were erected and stem cast and cured.

Design was more complicated than for simple spans because the unit was indeterminate forlive load. Accurate determination of the effects of time-dependent deformations on thecontinuous unit was not possible with design methods available at the time. “Design ofContinuous Highway Bridges with Precast, Prestressed Concrete Girders,”34 provided arational method for computing shrinkage and creep effects, but it was complicated forproduction design and subject to the usual scatter of prestressed beam camber and deflectionbehavior. Computer program PSTRS 13 was developed on a research project documented in“Automated Design of Continuous Bridges with Precast Prestressed Concrete Beams,”35 butits treatment of creep and shrinkage was unsatisfactory. By the time this program wasdebugged and updated for production, the use of prestressed beams continuous for live loadwas being abandoned.

The Waco District has used live load continuity for several bridges in excess of 300 ft. inlength with the bridge ends fixed by doweling the bridge slab to the approach slab and theapproach slab to the abutment wing walls. They are performing reasonably well.

The Bridge Design Section has designed a few such units in the 700 ft. range of continuouslength. This can be satisfactory if proper provision is made for expansion at the ends of theunit. Improper provision for expansion will lead to distress.

Fabrication Issues

Fabrication problems increased because of the extra hardware to be installed and allowed toprotrude from the beam end. The large-diameter bars had to be offset in adjacent beams toavoid fouling. This was a source of error that often resulted in fouling of the bars going


I-Beam Spans


unnoticed until the beams were erected. The project was delayed while the bars in one beamwere cut off and drilled and grouted in the proper position.

Construction Issues

Construction problems involved the diaphragm between beams. It was soon evident that thecontact surfaces of beam sides and diaphragm required insulation to prevent future beam endrotation from spalling the diaphragm. Inverted tee caps added a measure of awkwardness tothe construction schedules, because the stems could not be cast until the beams were erected.Segmental deck placement was more restrictive than for simple spans.

Maintenance Issues

Maintenance has not been extensive, but several unsightly conditions exist because ofinability to predict and accommodate time-dependent deformations of the beams. Beamssubject to significant camber growth were allowed to rotate with respect to the diaphragmcausing cosmetic spalling. Even after insulation of the interface was emphasized, severalseparation structures on a major highway bypass developed the symptoms before the projectwas opened to traffic. This problem was not manifest with inverted tee caps, nor does itoccur on all regular caps. One inverted tee unit has been given a concrete overlay because ofsevere slab cracking over the bent. The cracking appeared to be related to restraint of thelarge-diameter continuity bars on the plastic deformations of the concrete as the placementprogressed.

Current Status

The Bridge Design Section has decided that the disadvantages outweigh the advantages ofdesigning continuous for live load.

Design of precast prestressed concrete beam structures continuous for live load is notrecommended. Recent research36 reported that anywhere between 0 and 100 percentcontinuity for live load may be obtained, depending on the age of the beams when the slab iscast; also, performance is not influenced by the presence or absence of a positive momentconnection over interior bents. Without a positive moment connection, a continuous for liveload unit would be similar to a continuous deck only unit, which is currently recommended.


Although not recommended for new construction, the following suggestions are given forconsideration if there is an appropriate occasion to use this structure type:

♦ Use beams that are easily capable of the span length and beam spacing used. Try toavoid extreme camber after dead load deflection, which may lead to excessive cambergrowth.

♦ Rectangular caps may be used with a large interior diaphragm for continuityconnection. Details are shown in Figure 7-34.


I-Beam Spans


♦ Inverted tee caps are recommended. More effective positive moment connection can beachieved by extending the large anchorage bars into the stem. Aesthetics can beimproved by casting the stem flush with the outside beam shape. Cap shoring should besecurely in place until the stem is cast and cured. Figure 7-35 shows recommendeddetails.

♦ Recommended details for positive moment reinforcing of both types is shown inFigure 7-36.

♦ Design without consideration of creep and shrinkage is recommended.

♦ Deck slab reinforcement over the negative moment regions should be sized to resistultimate LL+I moment.

♦ Grade 60 reinforcing steel and Class S (currently 4,000 psi) cast-in-place concrete maybe used.

♦ Concrete strength in the beam may be controlled by the top stress at mid-span or thebottom stress over the supports.

♦ The stress range in mild steel reinforcing shall not exceed 21 ksi.


I-Beam Spans


Figure 7-34. Prestressed Concrete Beams Continuous for Live Load – Rectangular Cap(Online users can click here to view this illustration in PDF.)


I-Beam Spans


Figure 7-35. Prestressed Concrete Beams Continuous for Live Load – Inverted Tee Cap(Online users can click here to view this illustration in PDF.)


I-Beam Spans


Figure 7-36. Prestressed Concrete Beams Continuous for Live Load – Positive MomentConnections (Online users can click here to view this illustration in PDF.)

Chapter 7 — Superstructure Design Section 24 — Prestressed U-Beam Spans


Section 24 Prestressed U-Beam Spans

Background

The Bridge Division began development of the Texas precast concrete U-beam in the mid-1980s out of a desire to create an aesthetic, economic alternative to the much-usedAASHTO Type IV and Texas Type C precast concrete I-beams. The underlying premise inthe development of the U-beam was not to replace these concrete I-shapes, but to givedistricts the option of obtaining a different aesthetic bridge appearance with the economyand ease of precast construction.

The concept for the shape of the Texas U-beam began with a preference for a trapezoidalshaped beam. At the time, the Louisiana Department of Transportation used an open-toptrapezoidal beam that required a collapsible interior void form to fabricate. As a result oftheir input, as well as from precast concrete beam fabricators, the Texas U-beam was shapedto allow removal of the interior tub form without the need for a collapsible form. Additionaldetails for the beam were also generated with input from beam manufacturers as well as in-house experience with concrete I-beam construction details. The result of these effortsgenerated details for two sizes of U-beams: the U40 beam, which is 40 in. deep and 89 in.wide at the top, and the U54 beam, which is 54 in. deep and 96 in. wide at the top. Inaddition, the standard drawings contain details consistent with concrete I-beam construction.

Much has been learned since the first U-beam project was constructed in Houston in 1993.Because of the physical size of these beams, special consideration must be given to issuessuch as the haunch of the slab across the top width of the U-beam and the framing of thesebeams in flaring sections, curves, or long, narrow spans. Even secondary issues such asdrain details might prove to be difficult if not considered early in the design process.However, the Bridge Division has seen an increase in the number of bridge projects in Texasusing the U-beam. Popularity of the beam has even progressed to other states such asFlorida where the Florida Department of Transportation has developed an identical U-beamfor use in that state.

Preliminary bridge costs show U-beams to be a modest increase in cost over their concrete I-beam counterparts. Thus, it appears that U-beams are a viable precast concrete beam optionfor bridge projects in which aesthetic issues are deemed important.

Current Status

Complete sets of standard drawings for the Texas U-beam in English and metric units can bedownloaded from the TxDOT web site (http://www.dot.state.tx.us/). A checklist to assist inthe review of U-beam shop plans is also available from the Bridge Division. Standard cross-section dimensions, reinforcing, and the grid for possible strand locations for both the U40and U54 are shown in Figure 7-37.




With normal beam concrete strengths, the recommended economical span length limit is 100ft. for U40s and 120 ft. for U54s. These span lengths provide for a more efficient use of thenumber of beams in a given span. However, the maximum span lengths are approximately110 ft. for U40s and 130 ft. for U54s. Appendix A contains a table of U-beam spacingsversus span lengths for help in the preliminary layout of U-beams.

Slab details should show a cast-in-place slab with precast concrete panels since the standarddrawings are set up to work with this option. A full depth cast-in-place deck with permanentmetal deck forms is the optional design and the Permanent Metal Deck Form (PMDF(U))standard sheets have notes addressing the required slab reinforcing.

Use thickened slab ends at all expansion joints with non-inverted tee bents. See theMiscellaneous Slab Details (UBMS) standard sheets for details on thickened slab ends.

Do not show a detailed bill of reinforcing steel on production drawings. Instead, show atable of bar designations with sizes used in the slab as is currently done with I-beamstructures. In addition, show a table of estimated quantities with the total reinforcing steelbased on 3.7 lbs/sf of bridge deck. This quantity includes the extra slab steel required overinverted-tee bents and in thickened slab ends.

If inverted-tee caps are used and are sloped to match the sloping face of the U-beam, use a4:1 slope normal to the centerline of the bent. Avoid trying to take into account the actualcross-slope of the U-beams framing into the bent as this potentially complicates constructionof the bent cap. It is also suggested to extend the ends of the inverted-tee bents about 6 in.past the bottom edge of the exterior U-beam. This extension allows for a more definedbreak between cap and beams, especially since it is virtually impossible for the contractor toset the beams perfectly in line with the end of the cap. In addition, the Miscellaneous SlabDetails for Inverted-Tee Bents (UBMST) standard sheets show overhang details using thisconfiguration over the inverted-tee bent caps.



Figure 7-37. Standard Prestressed Concrete U-Beam Strand Pattern Constraints andSection Properties (See the following table of U-beam dimensions and section properties.Online users can click here to view this illustration in PDF.)

U-Beam Dimensions and Section PropertiesBeamType

Cin

Din

Ein

Fin

Gin

Hin

Jin

Kin

Ytin

Ybin

Areain2

Iin4

Weightplf

U40 89 40 33.25 57.5 27.0 10.125 8.375 17.0 23.66 16.30 979.9 183,108 1,021U54 96 54 47.25 64.5 30.5 24.125 11.875 20.5 31.58 22.36 1120.0 403,020 1,167



Slabs

♦ As with prestressed I-beam bridges, an 8 in. interior slab thickness should be usedwhenever possible. Thinner slabs should be used only if fewer beams will be requiredor if sufficient reduction in beam concrete strengths can be obtained. The minimumslab thickness will be 7 1/2 in.

♦ Virtually all projects use the continuous deck only concept. See Section 1, One-WayDeck Slabs on Stringers.

♦ The cast-in-place portion of slab contains Grade 60 reinforcement with #5 bars spacedat 6 in. in the transverse direction and at 9 in. in the longitudinal direction.

♦ For overhangs, use either an 8 in. thick normal overhang or a sloped overhang wherethe 8 in. dimension is applied at the edge of slab. The standard overhang dimension is 6ft.-9 in. measured from the centerline of the bottom of the exterior U-beam to the edgeof slab. For overhangs in excess of 7 ft.-3 in., the outside web-to-bottom flange joint ofthe exterior U-beam needs to be checked for adequacy under construction loads.Consideration should be given to using a normal overhang when conditions are presentthat might make the sloped overhang unsightly and/or difficult and expensive toconstruct. For the sloped overhang condition, the slope of the bottom face of theoverhang may vary significantly when used with curved slab edges primarily due to theoverhang distance varying along the length of the exterior U-beam. However, on astraight bridge slab edge, the slope of the bottom face of the overhang will vary due toonly the vertical curvature of the roadway surface and to the camber and dead loaddeflection of the exterior U-beam and, thereby, have a more pleasing appearance.

Beam Designs

♦ Normal concrete strengths for beams should be limited to f 'ci = 6,500 psi and f 'c =8,500 psi.

♦ Prestressing strands typically are 1/2 in. diameter 270 ksi low-relaxation strands.Although 0.6 in. diameter strand is available for use with the standard grid locations, itshould be used only when necessary.

♦ For interior U-beams, use a live load distribution factor of S/11 per truck/lane with aminimum value of 0.9 (S is the interior beam spacing measured between beamcenterlines).

♦ For exterior U-beams, use a live load distribution factor of 0.9 x Sext/11 per truck/lanewith no minimum value. (Sext = 1/2 × interior beam spacing + distance from centerlineexterior beam to edge of slab)On live load distribution factor (lldf) equations: the interior U-beam formula, S/11, andthe exterior U-beam formula, 0.9 × Sext/11, were decided upon for use by the BridgeDivision after studying several methods of lldf calculation including the formulas forspread boxes recommended by AASHTO. These equations generate values 3 to 10percent less than the lldf calculations made using simple beam distribution with thecenterline of beam flange as the supports for a given beam spacing, but are significantlymore than those values from the AASHTO formulas when they apply. Using thismethod, the minimum lldf value for an interior U-beam would be 1.0. However, the



lower limit of 0.9 for the interior beam represents the thought that some additionaldistribution of truck/lane loads is occurring between closely spaced U-beams.

♦ Overlay should be included if the district plans to immediately overlay the structureafter construction, or it can be included at the discretion of the designer. However,including overlay on the design of U-beams can significantly limit their ability to spanthe longer span lengths.

♦ Use 2/3 of the rail dead load on the exterior beam and 1/3 of the rail dead load on theadjacent interior beam.

♦ Each U-beam has two interior diaphragms at a maximum average thickness of 13 in.They are located as close as 10 ft. from midspan of the beam. Each diaphragm shouldbe accounted for as a 2 kip load for U40s and 3 kip load for U54s on the non-compositesection.

♦ Stresses at the ends of the beam are controlled with the use of debonding. Drapedstrands are not permitted in U-beams. The maximum amount of debonding is limited to75 percent of the strands per row and per section. The maximum debonded length is thelesser of the following:

1. half-span length minus the maximum development length as specified in the 1996AASHTO Standard Specifications for Highway Bridges, Section 9.28,

2. 0.2 times the span length, or

3. 15.0 ft.

♦ Grouping of U-beam designs are at the discretion of the designer. However, no exteriorU-beam shall have less carrying capacity than that of an interior U-beam of equallength. If the designer chooses to group beams, a general rule of thumb is to groupbeams with up to a four-strand difference.

Bearing Conditions

♦ Bearing pads shall be designed according to current TxDOT criteria for size andthickness, and shall conform to details shown on the Beam End and Bearing Details(UBB) sheets. U-beams rest on a three bearing pad system with two pads on the backstation beam end and one pad on the forward station beam end. The UBB sheets alsoshow standard distances to centerline of bearings and ends of beams for abutments,inverted-tee bents, and conventional bents.

♦ A left and right bearing seat elevation is given for each U-beam bearing seat location.Bearing seats for U-beams are level perpendicular to the centerline of bent but slopeuniformly between the left and right bearing seat elevations. This allows the bearingpads to taper in one direction.

♦ The designer should include a Bearing Pad Taper Report sheet in the plans thatsummarizes bearing pad tapers to be used by the fabricator. Appendix A containsinformation on the calculation of bearing pad tapers for U-beams.

Beam Framing

♦ U-beams are not vertical but are rotated to accommodate the average cross-slope of agiven span. As a result, the depth of slab haunch at the left and right top edges of the



beam may differ. Special attention should be given to these beams in calculating thehaunch values.

♦ The preferred method for framing U-beam centerlines is at the top of the beam. Thisprevents the spacing at the top of the beam from varying due to cross-slope of the beamand, thus, formwork dimensions for the slab are simplified for construction. Beamspacings shown on the span details should be noted as being at the top of the beam,while beam spacings shown on the substructure details need to take into account thehorizontal offset between the centerlines at the top and bottom of the beam.

♦ The alternate method for framing U-beam centerlines is at the bottom of the beam. Thismethod allows the U-beams to be framed as a vertical member whereby the beamspacings dimensioned on the span details and/or beam layouts match the beam spacingsshown on the substructure details. However, if this method is used, the designer shouldcall attention to the variable beam spacing at the top of the beams in the plans. Arecommended construction note to include on the span details is “Beam spacing shownis measured at bottom of beam. Beam spacing at top of beam may vary due to cross-slope of U-beams.”The Bridge Division currently uses the Roadway Design System (RDS) program toframe U-beams. The latest version of RDS frames U-beams using the alternate methodmentioned above. However, the Bridge Division is currently seeking alternative bridgegeometry software that will frame U-beams using the preferred method. The RDSmanual includes information on three framing options specifically written for U-beams:Options 20, 21, and 22. These framing options help the designer calculate accurate slabhaunch values, bearing seat elevations, and bearing pad taper reports for U-beams usingthe alternate method.

♦ Use the same minimum haunch value for all U-beams in a given span when reasonableto do so.

♦ Left and right bearing seat elevations are located at the intersection of the edges ofbearing seats with the centerline bearings. When calculating these elevations for eachbeam seat, be careful to apply the appropriate deduct at that elevation point, that is, theminimum deduct at the correct elevation point and the maximum deduct at the otherelevation point. Typically, the minimum deduct and maximum deduct are each appliedat diagonally opposite corners of a beam in plan view.See Appendix A for information on calculating U-beam slab haunches. The informationis tailored for use with RDS, but the principles behind the method remain the same.

Restraining Superstructure Lateral Movement

♦ Slab dowels are used to provide lateral restraint when constructing U-beams withinverted-tee bents. These dowels are located at the top of the inverted-tee stem and arein a slotted pipe to allow for expansion and contraction of the unit. Typically, onedowel is placed at the centerline of every beam 1 ft. from the centerline of bent. Slabdowels only need to be placed on one side of the centerline of bent. The criteria forlocating slab dowels within units are similar to the method used for locating dowelswithin concrete I-beam units.



♦ Concrete shear keys (or some other form of lateral restraint) between U-beams arerecommended for superelevated cross-sections on curves or on cross-sections sloping inone direction on straight roadways as follows:

1. on abutment caps at the ends of simple spans

2. on abutment caps at the end of units if the first interior bent does not have slabdowels or other lateral restraints

3. on rectangular bent caps

Shear keys are not required when using slab dowels with inverted-tee bents. Thedesigner has discretion of the placement of shear keys between U-beams. However,consideration should be given to the transverse expansion of the slab, particularly withwide structures, in locating the shear keys. A suggested rule of thumb is to locate shearkeys no further than 40 ft. from either side of the structure centerline.

♦ Concrete shear keys are typically poured 5 in. above the bottom of the U-beam. Theyare designed as pedestals and are poured after the beams have been erected.Bituminous fiber material can be used as the bond breaker at the beam/shear keyinterface.

Chapter 7 — Superstructure Design Section 25 — Rolled Steel I-Beam Spans


Section 25 Rolled Steel I-Beam Spans

Background

Structural steel I-beams have been rolled in the U.S. since about 188737 I-beams made ofwrought iron were available for some 16 years previous. The earliest standard drawings inthe Texas Bridge Division file were prepared by the federal Bureau of Public Roads in 1917.The state produced its own standard drawing in 1919 based on designs by Percy V.Pennybacker, who was influential in steel design and construction for many years.

The earliest simple I-beam spans had timber decks, but reinforced concrete decks soonbecame prevalent. Span length capabilities were first increased by use of cantilever drop-inType 3 span units. Some of the advantages of continuity could be obtained without thestructure being statically indeterminate. Hinges were notched beam seats with bearings firstand pin and hangers later. By the early 1940s, continuous units with riveted splices werebeing designed. Simple spans still retained popularity because of simpler construction.

Then, in the late 1940s, Texas began to develop the capability of quality field welding ofbeam splices and framing members. Under the guidance of Percy Pennybacker, a program ofeducation and emphasis on weld quality was undertaken among designers, weldingcontractors, and construction inspectors. The Bridge Division hired several welders,instructed them on the basics of metallurgy, and assigned one to each construction project onwhich field welding was required. Their purpose was to observe the welders at work and dowhatever was necessary to get good welds. In the process, the district field personnel learnedwhat to watch on future work. These Bridge Division welding inspectors were generallyhard nosed by virtue of their former ability to excel in a competitive trade, and apparentlythey were well motivated to carry out the wishes of the department. All of this plus a liberaluse of radiographic testing (RT) can be credited for Texas’ long usage of field weldingwithout significant fatigue or fracture problems.

With the simplified splice details allowed by welding, simple spans became a thing of thepast. Continuous units were used extensively in the 1950s. Almost any geometriccombination of skew, curvature, and flare could be handled with straight beams betweensplices. Heat curving or heat cambering of I-beams was seldom successful. I-beams, ingeneral, ceased to be economical in the early 1960s. They succumbed to rising steel pricesand a new and vital prestressed beam industry in Texas. Standard drawings were maintainedthrough the 1960s but their use was minimal.

Various types of structural steel specified by the Texas Highway Department and AASHTOwith allowable design stress in flexural tension are shown in the "Chronology of AASHTOSpecification Requirements. . . " table.

The development of standard I-beam details is outlined in "Chronology of Simple Steel I-Beam Standards" table.



There has been a slight resurgence in the use of steel I-beam spans. Their cost iscomparable to concrete box beams, and steel is much easier to adapt to severe geometricconstraints.

Chronology of AASHTO Specification Requirementsfor Structural Steel I-Beam Era

AASHTOSpecification

ASTMSpecification

♦♦♦♦ Maximum AllowableBending Stress (ksi)

1918(THD)

Nonefy≥30 ksi

16

1926(THD)

A7 16

1931 A7 16LL 24DL1935 A7 181941 A7 (1) 181944 A7

A94 (2)A8 (3)

182430

1949 A7A242 (4)A94A8

1822, 24, 27

2430

1953 A7A242A94A8

1822, 24, 27

2430

1957 A7, A373 (5)A242A94A8

1822, 24, 27

2430

1996 A709, Gr. 36A709, Gr. 50A709, Gr. 50W

202727

(1) Carbon(2) Silicon(3) Nickel(4) Low alloy(5) Weldable carbon♦ Allowable compressive stress often blow 0.55 Fy



Chronology of Simple Steel I-Beam Standards

SeriesYear

DesignedLiveLoad

Live LoadDistribution Factor

Allowable Stress(psi)

1I-Beam bridges andculverts

1919 15T Rolleror 100 psf 0.5

S 16,000

I-24. . . 1929 15T Rolleror 100 psf 5.4

S 16,000

IB-22. . . 1930 15T Rolleror 100 psf 5.4

S 16,000

I-32. . . 1932 2-15T Trucks5.4

S 16,000

IL-24. . . 1937 1-10T Truck5.4

S 16,000

IL-2. . . 1937 H155.4

S 16,000

I-18. . . 1946 H200.5

S 18,000

I-9. . . 1946 H200.5

S 18,000

FI-9. . . 1950 H100.5

S 18,000

2Is-24(S16) 1960 H20S165.5

S 18,000

2Is-28(20) 1960 H205.5

S 18,000

1 . Adapted from Bureau of Public Roads and Rural Engineering Designs2 . For interstate highway crossovers

Current Status

The use of rolled steel I-beams by TxDOT is usually because of section depth limitations,severe skews, or other geometric anomalies.


The following suggestions are offered for the design of steel I-beam spans or units:


♦ Simple spans units with continuous slabs are preferred.

♦ Cover plates should not be used.Welded cover plates are highly susceptible to fatigue problems.

♦ Composite action should be assumed in dead load positive moment sections. Shearconnection should be provided, according to the specification requirements for



“Fatigue” at 500,000 cycles, at 24 in. maximum spacing of stud groups. A nominalnumber of additional connectors should be spaced at 6 in. at the span ends. A check forultimate strength is not required if service load design is used.These are local specification interpretations intended to simplify and standardize thedesign of shear connectors.

♦ Beam splices should be minimized and, if used, should be welded. Field splices shouldbe welded and optional bolted designs provided. Beams, if spliced, should be fromwithin the same rolling family.

Recommendations of the Structural Steel Quality Council should be followed in the designand detailing of rolled I-beams and can be found at the web site www.steelbridge.org.

Chapter 7 — Superstructure Design Section 26 — Steel Plate Girder Spans


Section 26 Steel Plate Girder Spans

Background

Since the 1930s, I-shaped plate girders have been used to span beyond the range of rolledbeams. Originally, girders were fabricated by riveting flange angles to a web plate andadding cover plates top and bottom. The most common configuration was two girdersconnected by transverse floorbeams with rolled beam stringers parallel to the girders, toppedby a one-way deck slab. Quasi-continuity was occasionally provided with pin and hangerjoints, but usually full continuity was accomplished with riveted splices. The last of these ona highway bridge was constructed in the early 1950s.

Various types of structural steel specified by TxDOT and AASHTO are shown in the"Chronology of AASHTO Specification Requirements" table . Welded girders were usedfirst on tangent alignment at about 8 ft. spacing topped with a one-way deck slab. Detailsimproved as fabrication and construction experience increased. Design, detailing, andfabrication of steel girders became much simpler when welding was accepted as a qualityconnection technique. Flange angles were no longer needed since the web plate could bewelded directly to the flange plates. Flange splices could be butt welded in the shop andfield splices butt welded on the job. There were no more rivet spacings and splice plates todesign, detail, and fabricate. Structural steel weights for typical tangent, closely spaced I-girder units are shown in the “Structural Steel Weights for Tangent, Closely Spaced I-Girders” table .



Chronology of AASHTO Specification Requirementsfor Structural Steel Plate Girders Era

AASHTOSpecification

ASTMSpecification

♦♦♦♦ Allowable Fbt(ksi)

1953 A7 (1)A242 (2)A94 (3)A8 (4)

1822, 24, 27

2430

1957 A7, A373 (5)A242A94A8

1822, 24, 27

2430

1961 A7, A373A242, A440, A441A94 (6) (7)A8

1822, 24, 27

2430

1965 A36 (8)A441 > 4 in.A242, A440, A441

2022

23, 25, 271969 A36

A441 > 4 in.A242, A440, A441, A588 (9)A514/517 (10)

2022

23, 25, 2749, 55

1973 A36A441 > 4 in.A242, A440, A441, A588A572A514/517 (10)

2022

23, 25, 2723-3649, 55

1977 A36A441 > 4 in.A242, A440, A441, A588A514/517

2022

23, 25, 2749, 55

1983 A36A572, A588A514/517

2027

N.A.1989 A709, Gr. 36

A709, Gr. 50(W) (11)A709, Gr. 70(W)A709, Gr. 100(W)

2027

N.A.N.A.

1996 A709, Gr. 36A709, Gr. 50A709, Gr. 50W

202727

Proposed Specifications A709, Gr. HPS70W 38



(1) Carbon(2) Low alloy(3) Silicon(4) Nickel(5) Weldable carbon(6) Low alloy rivet steel(7) Manganese vanadium low alloy(8) Weldable high yield carbon(9) Weathering low alloy(10) Quenched and tempered(11) Covers all and includes AASHTO Charpy impact requirements♦ Allowable compressive stress often below 0.55 Fy



Structural Steel Weights for Tangent, Closely Spaced I-GirdersSpan Lengths

(ft)Deck Width

(ft)Structural Steel Weight

(lbs/ft2)86-104-108 (1) 42 32

120-150-120 50 35140-180-140 44 49

150-180-180-150 58 43180-240-180 44 53

198-270-198 (2) 31 59(1) Constant depth(2) Variable depthWeights are from actual designs.

When geometric requirements began to extend span lengths and complicate framing, curvedI-girders became appropriate. Structural analysis was difficult, and some of the earlierdesigns were based on engineering judgment. Curved units were always heavier because oftorsional effects and, with the additional complications of fabrication and erection, structurecosts were quite high. There was no economical alternative, since Texas had not developed aconcrete box girder capability. Structural steel weights for typical curved I-girders units areshown in the following “Structural Steel Weights for Curved, Constant Depth I-Girders”table.

After some design studies for long span steel girders it became evident that steel weightcould be saved by increasing girder spacing. Spacings from 12 to 27 ft. with two-way slabsand no stringers were used on tangent girders spanning from 175 to 480 ft. Texas’ firsthybrid girder was constructed in this manner, as well as a few intracoastal canal bridges andcrossings of the now defunct navigation channel of the Trinity River. Variable depth girderswere almost always used with this system. Structural steel weights for typical widely spacedI-Girder units are shown in the following “Structural Steel Weights for Tangent, WidelySpaced I-Girders” table.



Structural Steel Weights for Curved, Constant Depth I-GirdersSpan Lengths

(ft)Deck Width

(ft)Radius of Curve


(lbs/ft2)96-90-96-90 27 160 3968-123-117 42 440 42

108-134-108 44 440 4698-181-84 42 440 61175-175 28 570 55

180-194-166 28 640 53210-210 28 670 73

193-208-204 38 880 61121-181-95-112 32 960 54

92-145-145-145-92 42 1150 40155-155-155-155 42 1150 63160-200-200-160 28 1270 48

124-248-124 44 1270 66122-115-122 40 1430 29

93-113-113-104 44 2000 30Weights are from actual designs and may be subject to variations according to the design methodused.

Structural Steel Weights for Tangent, Widely Spaced I-GirdersSpan Lengths

(ft)Girder Spacing

(ft)Deck Width


(lbs/ft2)150-190-190-150 26 36 26150-190-190-150 17 44 27

175-200-175 16 40 35200-250-250-200 14 37 41

100-280-200 20 50 44200-290-200 20 50 40220-310-220 17 44 46220-310-220 27 68 53220-310-220 28 70 46240-340-240 24 62 52320-480-320 23 58 79

Weights are from actual designs.

Steel structures have a history of fatigue cracking and brittle fracture. Some weld detailshave been shown by numerous laboratory tests to be highly susceptible to fatigue crackinitiation in bridge girders and I-beams. Growth of the crack and possible subsequent brittlefracture are influenced by the toughness of the steel. Crack initiation is independent of steelstrength, but toughness is a function of strength, ingredients, method of manufacture,



thickness of plate, and temperature. Quenched and tempered (Q&T) steel was notorious forbrittle fracture and was later disallowed for flexural members. Texas has only one bridgecontaining this material in a flexural member. In recent years a new high performance steel,HPS70W, has been made available. There is a quenched and tempered version of this steelthat has high toughness characteristics. Although somewhat more costly than conventionalgrade 50W steel, its use is now encouraged where applicable.

For several years, the steel industry, AASHTO, and FHWA argued about the level oftoughness required for bridges and how to ensure that the steel used has the toughnessrequired. There is now an acceptable program of Charpy impact testing, varyinggeographically with service temperature and amplified for structures considered “fracturecritical.”

Fabrication Issues

Weld details are very critical to good performance and also to economical fabrication.Improvement of weld details was a continuous process for several years.

Thin web plates are a source of concern. The specification allowed very thin web plates, andoverall steel weight was minimized by using the thinnest web possible. Web plates 0.3125in. thick were impossible to fabricate without significant cupping between stiffeners. Webplates 0.375 in. thick were barely tolerable. Distortion due to weld shrinkage is a persistentproblem with thin web plates.

For curved girders, the main problem in fabrication is maintaining the proper horizontalcurvature. The specifications allow heat curving, but cutting flange plates to the curvature ispreferred by some fabricators.

For widely spaced girders, fabrication is complicated by variable web depth, longitudinalstiffeners, braced floor beams, and lateral bracing. Shipping problems are more significantbecause of excessive depth of the negative moment sections and length of the positivemoment sections. The 480 ft. span girders were 20 ft. deep over the supports, and ahorizontal field splice was required in the web plate so the pieces could be shipped. Thehorizontal splice was bolted, but all vertical field splices were welded.

Erection Issues

Girder sections tend to be unstable during erection if the top flange plate is too narrow.Composite design tends to make the top flange narrow. Erection procedures, includingfalsework and girder support while welding or bolting splices, require close coordinationwith the contractor.

Erection is especially critical for curved girder units. Accurate analysis of deflections andstresses during erection stages is very complicated and time consuming. The tendency is torely on judgment until trouble occurs. Except for a few notable mistakes, Texas hasmanaged to get its curved girders erected without significant mishap. Close coordinationbetween field and design personnel is vital.



Erection problems are increased by the size and weight of the girders. Most of TxDOT’slarge, widely spaced girders are in three-span continuous units. Falsework was usuallyprovided at the end span splices at the inflection points. Positive moment pieces that weretoo long to ship in one piece were usually spliced on the ground and lifted into place. Endspan splices over the falsework were welded. Usually, the center span section was hungfrom the two cantilevered ends of the haunched section by temporary shear plates andaligned with clamps. Once started, splice welding was continuous until at least half of thesplice in each flange was complete, to prevent shrinkage cracking. One such unit,completely erected but without a deck slab, was observed to develop harmonic response to a60 mph norther. Amplitude of the center span deflection ranges was estimated to be 3 ft.Subsequent inspection revealed no evidence of damage to the welded connections.

Maintenance Issues

There have been a few cases of cracking due to out-of-plane bending. This usually occurson heavily skewed units with diaphragms perpendicular to the girders. Deck cracking issometimes more severe due to long continuous slab placement. Painting has been unreliable.In 1978 after serious weld flaws were discovered in a 1957 vintage girder, ultrasonicinspection was performed on girder units constructed during that time frame. Numerouswelds were found to be outside the current limits of acceptable ultrasonic performance, butno weld cracking was observed. All of these bridges were redundant so no action was takento repair the questionable welds. One instance of brittle fracture was reported in 1977. Itstarted at a shop splice in the bottom flange of a girder constructed on a county road andinspected by county personnel in the early 1950s. When the flange crack was discovered, itwas 0.75 in. wide. Texas experience does not appear to justify the severity of the currentAASHTO fatigue requirements. There are more than 3,000 steel girder bridges on the statehighway network. At least 45 percent of these were constructed after 1951 using all weldeddetails. The lack of fatigue and fracture problems has been attributed to good weld qualitycontrol and mild climate.

Except for paint, there have been few maintenance problems with curved girders. Deckcracking may be a little more prevalent on curved girder units but not to the extent thatcauses extra maintenance procedures. In one instance after a design error forced the needfor false bents and composite action for slab dead load, cracking in the negative momentarea was particularly prominent. The structure was built in the late 1970s and is servingtraffic well.

Paint has not been a maintenance problem for widely spaced girders since most werefabricated of weathering steel and left bare. One of these, located about 2 miles from theGulf of Mexico, is subject to controversy. The Bridge Design Section considers it to beweathering, but the Construction Division says it is rusting. A recent inspection discoveredthe possibility that loss of section is continuing under the bridge in the variable depthsection. The theory is that the seclusion of this area discourages wind drying of the salt-laden atmosphere.

The structures with problems previously mentioned are still carrying traffic. The weatheringsteel bridge near the coast has not been painted. Although paint and painting techniqueshave been improved somewhat in recent years, the Bridge Division encourages the use of



unpainted weathering steel except in northern urban regions of the state where de-icing saltsare used heavily.

Current Status

Steel I-shaped plate girders should be used when the span lengths and/or horizontalcurvatures exceed the capabilities of prestressed beams. Closely spaced constant depthgirders should be used for tangent spans up to about 300 ft. and for all horizontally curvedunits. Widely spaced variable depth girders should be considered for use on tangent spansgreater than 300 ft. where vertical clearance is not critical. Unpainted weathering steel ispreferred for all girders not subject to continuous moisture or in northern urban areas wherede-icing salts are used.


Girder design in general, including tangent constant depth girders, will be discussed first,followed by specific considerations for curved girders and widely spaced variable depthgirders.

General - Tangent - Constant Depth - Closely Spaced

The 1989 AASHTO Specifications announced new American Society of Testing andMeasurement (ASTM) steel specification as follows:

♦ A709, Grade 36 was A 36 (Grade 250)

♦ A709, Grade 50 was A 572 (Grade 345)

♦ A709, Grade 50W was A 588 (Grade 345W)

For girders that are to remain unpainted, weathering steel is required. Available grades ofweathering steel are A709 Grades 50W and HPS70W. If the girders are to be painted,several grades of steel may be used, including weathering steel. These grades are A709Grades 36, 50, 50W, and HPS70W. Note, Grade 50W should not be required for paintedgirders.

With little to no cost advantage over Grade 50 or 50W steel, the use of Grade 36 steel shouldbe minimal; therefore, the use of hybrid girders with Grade 36 webs is discouraged. IfHPS70W is used, it should be in a hybrid girder with a Grade 50W web if the girder is toremain unpainted or a Grade 50 web for painted girders.

Closely spaced girders are usually not considered fracture critical.

♦ Service load design is recommended.

♦ Specification loads and load distribution apply without modification.

♦ Preliminary analysis may be done with the help of influence lines such as published in“Ten-Division Influence Lines for Continuous Beams,” and “Moment Shears andReactions for Continuous Highway Bridges.”38 Final analysis can be done with the B-30 computer program.39



♦ Deck slab should be composite with girder in the dead load positive moment zonesonly. Shear connectors should be provided according to specification requirements for“Fatigue,” at 500,000 cycles. Spacing of stud groups should not exceed 24 in.“Additional Connectors to Develop Slab Stresses” should be spaced at 6 in. centers atthe contraflexure points and also at the ends of the unit. There should be no less thanfive rows spaced at 6 in. for this purpose. A check for ultimate strength is not required ifservice load design is used.

♦ Web plates should be no less than 0.5 in. thick because of fabrication problems.Additionally, they should be thick enough to eliminate the need for transverse stiffenersexcept within two times the web depth adjacent to the interior supports. Longitudinalstiffeners should not be used.

♦ The minimum flange plate width should be the length between field splices divided by80 for erection stability. The minimum thickness should be 0.75 in. and the maximumwidth 24 times the thickness. The flange thickness should not exceed 4 in.

♦ Shop flange splices shown on the details should be minimized. Usually one in thebottom flange either side of the maximum positive moment location and one in bothflanges either side of interior supports will suffice.

♦ Shop web splices should preferably not be shown on the details. Thickness changesconsidered appropriate could be made at the field splices.

♦ The following girder fabrication note should be placed on the details:Except at changes in section, shop flange and web splices in plate girders may belocated as desirable to optimize plate lengths and erection procedures, except thatsplices will not be allowed where a 40 ft. or less unspliced length would suffice; neitherwill tension flange splices be allowed within .05S either side of interior bearings, within.10S either side of the centerlines of interior spans, nor within the range between .30Sand .50S from the end bearings (S = length c/c bearing of span in which the splice ismade). The fabricator may lengthen thicker plates if the contractor approves any changein thickness at field splices. Flange and web splices shall be made by full penetrationgroove welds in accordance with the Item “Steel Structures.”

♦ Field splices should be located near the dead load inflection points in each span.

♦ Permissible field splices should be used to keep piece lengths less than 130 ft.

♦ In keeping with past practice, all field splices are to be shown in the plans as weldedsplices. A new requirement is to also design and detail bolted field splices as an optionto welded splices.

♦ It takes a lot of traffic to generate more than 2,500 average daily trucks in one direction.Unless a current Transportation, Planning and Programming Division traffic count nearthe proposed bridge indicates more than 2,500 combination and three-axle trucks ineach direction, Case II loading cycles should be used.

♦ Stress range should be limited to the allowable for Category C, details which cover studshear connectors and transverse stiffeners welded to the flange.

♦ Transverse stiffeners to which diaphragms are connected should be welded to the topand bottom flanges. This will minimize out-of-plane bending of the web.



♦ Bottom compression flanges are considered supported at the diaphragms and bearings.The allowable compressive stress shall conform to the AASHTO Specifications.

♦ For hybrid girders, the combination of high shear and bending should be checkedadjacent to interior supports.

Miscellaneous details are shown on standard drawings SPGD. Weld details are given inTxDOT construction specification Item 441 “Steel Structures” and in the AASHTO/AWSBridge Welding Code. Welding symbols are explained in the ANSI/AWS A2.4-93.

For bolted connections or splices it is recommended that ASTM A325 bolts be used with adesign allowable shear stress of 15 ksi (assuming 0.33 slip coefficient). Blast-cleaned fayingsurfaces are specified in construction specifications for System II paints.

Recommendations of the Texas Steel Quality Council should be followed in the design anddetailing of structural steel girders and can be found at the web site www.steelbridge.org.

Curved Girders - Constant Depth - Closely Spaced

The AASHTO Guide Specification for Horizontally Curved Highway Bridges40 givesmaximum central angles beyond which torsional effects must be considered in primarybending calculations. In the Bridge Design Section this guide is applied to the longest spanin a continuous unit. If the central angle is less than allowed by the guide, design of the unitis treated, in all respects, like a unit on tangent alignment. If not, it is subject to acomplicated procedure considering torsion and lateral flange bending. The guidespecification is then subjugated to local practice.

Material selection criteria are the same as for tangent girders and loads. Load distributionand allowable stress from the AASHTO Specifications for tangent girders are used except ashereinafter noted.

Service load design is recommended.

An approximate method of analysis has been developed using the V-Load concept41 and B-30 computer program, which is considered sufficiently accurate for most situations.

♦ Establish a preliminary design using B-30.

♦ Calculate V-Loads for non-composite dead load and also for any composite dead load.

♦ Find the maximum live load moment in each span and at each support.

♦ Compute separate loadings, concentrated at each diaphragm, that will approximatelyreproduce each maximum live load moment.

♦ Remove all dead load and live load from the B-30 input and make a separate B-30 runfor each concentrated loading configuration.

♦ Calculate V-Loads from the moments due to each loading.

♦ Apply the V-Loads as composite P-Loads in a separate B-30 run for each configuration.

♦ Select the maximum percentage increase in live load moment caused by the V-Loads;that is, the absolute maximum from the several live load/V-Load runs.



♦ Make a final B-30 run for each girder with dead loads and live load using the followingadjustments:

1. Apply non-composite dead load V-Loads as non-composite P-Loads to each girder.

2. Apply composite dead load V-Loads as composite P-Loads to each girder.

3. Increase the live load distribution factor by the percentage selected above, for theoutside girder.

4. Increase live load proportionally for other girders outside of the centerline of thesuperstructure.

5. Do not increase or decrease live load to girders inside of the centerline.

This method produces a final moment B-30 run for each girder that will usually suffice forstress ranges, shear connector spacing, and flange plate cut-off. The method is approximatebut conservative. Since the load case giving the largest percentage increase in live loadmoment is used to adjust the distribution factor, other areas of the outside girder will beoverdesigned. The amount of over design will usually be an insignificant portion of the totalstresses in the girder. Engineering judgment may suggest modifications to this procedure asthe design progresses. If desired the design may be made using the DESCUS I computeranalysis program.

Because of lateral bending stresses in the flanges due to non-uniform torsion, diaphragmspacing and flange plate width are significant design items. Lateral flange bending stressesmust be calculated by hand and added to the final stresses above. Since the top flange islaterally restrained by the slab against live load effects, it will usually be sufficientlyaccurate to calculate the maximum lateral flange bending stress in each size of bottomflange. Formulae for lateral flange bending are as follows:

( )bM

12hRdM

(Bimoment)Moment Bending Flange Lateral 2

==

where:M = Maximum moment (including V-Loads) within the flange plate size under considerationH = Girder web depthR = Radius of horizontal curvature of the girderd = Diaphragm spacing

2b

tw6M

Bending Lateral toDue Stress Tip Flange = (Conservatively applied at all locations along girder)

where:t = Flange thicknessw = Flange width

These formulas can be quickly applied to the appropriate moment from the final B-30 run toproduce an additional stress due to lateral flange bending, which can be added to thecombined DL/LL stresses and stress ranges, as necessary.

Careful attention to girder reactions at the ends of unit is appropriate because of uplifttendencies at the inside girder. Although theoretical movement of the girder ends is along achord to the curvature, expansion bearings are usually set square with each girder.



Lateral bracing is never required to resist wind forces in properly designed curved girders.Although research42 suggests that lateral bracing increases the resistance of the girder totorsion, it has not been used for that purpose by the Bridge Design Section.

Additional design considerations for curved girders are as follows:

♦ Diaphragm spacing should be less than 20 ft.

♦ Standard diaphragms must be checked against actual forces. Modifications to thestandard details are often required.

♦ Flange plate width should be no less than one-fourth of web depth. However, one-thirdof the web depth is preferred.

♦ Calculated bending stress plus lateral flange bending stresses should not exceed0.55 Fy.

♦ Live load deflection is calculated for the outside girder assuming only one lane loaded.The allowable live load deflection is one eight-hundredth of the span along the outsidegirder from bearing to bearing.

Miscellaneous details are shown on standard drawing SPGD. Under no circumstancesshould X diaphragms without top horizontal members be used. Diaphragm connectionstiffeners should always be welded to both girder flanges.

Standard drawing MEBR(S) contains some empirical controls on the erection of curvedgirders. Application of these controls will often require a false bent between field splices,which will be unacceptable if traffic must be maintained beneath the unit. Elimination of thisfalse bent requires temporary support of two or three girders until they can be made stableby diaphragm connection and/or partial splicing. Current policy requires the contractor toprove the adequacy of any erection procedure that does not conform to MEBR(S).

Tangent Girders - Variable Depth - Widely Spaced

Fabricators discourage the use of variable depth girders. Widely spaced girders with a two-way deck slab will usually be more economical than closely spaced girders but will requiremore depth. The recommended minimum span length for this type of construction is 300 ft.TxDOT’s longest constructed span with this framing system was 480 ft. (IH 45 over theTrinity River, south of Dallas). Widely spaced variable depth girders are not recommendedfor curved structures. Vertical clearance is usually critical, and variable depths are notconsidered aesthetic on a curve.

Material selection is similar to the other girder types. Flange plates will be large on longspan units. Four in. thickness should be considered the absolute maximum. Grade 50, 50W,or HPS70W steel should be used. Quenched and tempered Grade 100 should never be usedfor flexural members.

♦ Two-girder units are considered fracture critical and should be avoided. The additionalcost of fabrication according to the “Fracture Control Plan” is currently about 15percent.

♦ An example configuration of diaphragm floorbeams is shown on Figure 7-38.



♦ Current AASHTO design methods will seldom require lateral bracing. If lateral bracingis required it is recommended that l/r not exceed 140. This will usually producemembers with more than enough strength to withstand wind forces. Example bracingdetails are shown in Figure 7-39.

♦ Web thickness should be no less than 0.5 in. but longitudinal and transverse stiffenersmay be desirable to reduce weight, especially in the deeper negative moment zone.Since longitudinal stiffeners are potential Category E fatigue details, the recommendedconnection details from standard drawings SPGD are reproduced on Figure 7-40 foremphasis.

♦ Other suggestions for tangent, closely spaced, constant depth girders are also applicable.

♦ Example details of the compression flange connection at interior bearings are given inFigure 7-41.

♦ Bearing and expansion joints will be more specialized for the longer units. This will bediscussed in later sections.



Figure 7-38. Widely Spaced Girder Details (Online users can click here to view thisillustration in PDF.)



Figure 7-39. Example Lateral Bracing Details (Online users can click here to view thisillustration in PDF.)



Figure 7-40. Longitudinal Stiffener Details (Online users can click here to view thisillustration in PDF.)



Figure 7-41. Variable Depth Girder Details (Online users can click here to view thisillustration in PDF.)

Chapter 7 — Superstructure Design Section 27 — Trapezoidal Box Girders


Section 27 Trapezoidal Box Girders

Background

Except for the one unorthodox through box girder in Austin, TxDOT avoided the use ofsteel box girders until the late 1980s. However, girders with sloping webs and weatheringsteel have existed since 1968 on the road system of the Dallas-Fort Worth airport.

Beginning with the toll roads constructed by the Metropolitan Transit Authority in Houston,trapezoidal box girders became an acceptable structure type, especially in the urban areas.

The Houston District has now designed several such structures, consultants have designedsome, and the Bridge Design Section has even designed a few. Center spans of thesecontinuous units are usually well over 200 ft., being in the core region of an interchange, andthe girder sections are necessarily large and heavy, compared to I-girders. Fabrication ismore complicated, especially on a horizontal curve. Shipping and erection may requirespecial measures.

Current Status

Trapezoidal box girders are recommended in interchanges where aesthetics dictate the use ofU-beam approaches. Example details are shown in Figure 7-42.



Figure 7-42. Example Trapezoidal Steel Box Girder Details (Online users can click here toview this illustration in PDF.)


Material selection criteria should be the same as for conventional girders. A minimum 8 in.thick Class S concrete slab should be supported during placement with permanent metaldeck forms. Access holes into each girder should be provided near the end bearings.Lockable doors must cover each access hole. Crawl holes must be provided in the bearingstiffener/diaphragms at interior bents. In addition to the diaphragms within the box,temporary crossframes should be provided between girders to help the girders deflecttogether during slab placement. They may be removed after the slab has cured. Otherrecommendations are as follows:


♦ The B-30 computer program can be used for tangent girders and preliminary curvedgirder sections.

♦ DESCUS II is available and may be used for analysis of curved units.

♦ The minimum tension flange thickness should be 0.50 in.

♦ Field splices should be bolted, using 1 in. diameter, A325 bolts with a design allowableshear stress of 15 ksi. Faying surfaces should be sandblasted.



A long depth transition is desirable to meet the approach span depth. Lateral bracing withinthe box should be bolted to the top of the flange to mitigate fatigue problems and to avoidconnection difficulties to the sloping web beneath the top flange.

Every attempt should be made to use steel reinforced neoprene bearings. However, highload-multirotational bearings (usually pot bearings) may be required because of the largereactions. One or more bearing manufacturers should be contacted before bearing seatdimensions and connection details are finalized.

Spans with two box girders in the cross section are considered fracture critical, which adds aconsiderable amount to fabrication costs.

Recommendations of the Texas Steel Quality Council should be followed in the design anddetailing of trapezoidal box girders and can be found at the web site www.steelbridge.org.

1 “Standard Specifications for Highway Bridges,” American Association of State Highway and TransportationOfficials (AASHTO), Sixteenth Edition (1996).2 “A Study of Prestressed Panels and Composite Action in Concrete Bridges Made of Prestressed Beams,Prestressed Sub-Deck Panels and Cast-in-Place Decks,” Furr, H.L., and others, TTI, Reports 145-1, 2, 3, & 4F, 1970& 1972.“Behavior of Prestressed Panel Cast-in-Place Concrete Bridge Decks,” Bieschke, L.A., and R.E. Klingner, CTR,Report 303-1F, 1982.3 “A Discrete-Element Method of Analysis for Orthogonal Slab and Grid Bridge Floor Systems,” Panak, J.J., and H.Matlock, CFHR, Report 56-25, 1972.4 “A Discrete-Element Method of Analysis for Orthogonal Slab and Grid Bridge Floor Systems,” Panak, J.J., and H.Matlock, CFHR, Report 56-25, 1972.5 “Influence Surfaces of Elastic Plates,” Pucher, Adolf and Springer-Veriag Wein, New York, 1977.6 “Standard Specifications for Highway Bridges,” American Association of State Highway and TransportationOfficials (AASHTO), Sixteenth Edition (1996).7 “Behavior of Concrete Slab and Girder Bridges,” Leyendecker, E.V. and J.E. Breen, CFHR, Final Report 94-3F,1969.8 “Experimental Use of High Strength Reinforcing Steel” Newton, J.G. and L.G. Walker, Departmental Research,Final Report 25-1F, 1966.9 “Application of Transverse Prestressing to Bridge Decks,” Poston, R.W. and others, CTR, Reports 316-1, 316-2,and 316-3F, 1985.10 “Application of Transverse Prestressing to Bridge Decks,” Poston, R.W. and others, CTR, Reports 316-1, 316-2,and 316-3F, 1985.11 “The Analysis of Continuous Beams for Highway Bridges IV,” Georgia D.O.T., 1971, Enhanced by SDHPT.12 “Prestressed Concrete,” Guyon, Y., John Wiley and Sons, 1955.13 “A Computer Program to Analyze Beam Columns Under Moveable Loads,” Matlock, H. and T.P., Taylor, CTR,Report 56-5, 1968.14 “Design of Slender Non-Prismatic and Hollow Concrete Bridge Piers,” Poston, R.W., and others, CTR, FinalResearch Report 254F, 1983.15 “Bridge Design Practice,” State of California, Department of Transportation, Latest Revisions.16 “Guide Specifications for Design and Construction of Segmental Concrete Bridges,” AASHTO, 1989.17 “Prestressed Concrete,” Guyon, Y., John Wiley and Sons, 1955.18 “Guide Specifications for Design and Construction of Segmental Concrete Bridges,” AASHTO, 1989.19 “Tentative Standards for Prestressed Concrete Piles, Slabs, I-Beams and Box Beams for Bridges,” Joint AASHTOand PCI Committee Report, 1962.20 “Economical Precast Concrete Bridges,” Panak, J.J., Departmental Research, Final Report 226-1F, 1982.21 “Prestressed Concrete Girder Design,” Computer Program PSTRS 14, Automation Division, SDHPT, 1990.22 “Precast Prestressed Concrete Short Span Bridges – Spans to 100 Feet,” Prestressed Concrete Institute, 1975.23 "The Effect of Diaphragms in Prestressed Concrete Girder and Slab Bridges," S. Sengupta and J. E. Breen, Centerfor Highway Research, Report #158-1F, 1973.24 “Differential Camber in Prestressed Concrete Beams,” Jones, H.L. and H.L. Furr, TTI, Report 193-1F, 1977.



25 “Fatigue Behavior of Pretensioned Concrete Girders,” Overman, T.R. and others, CTR, Report 300-2F, 1984.26 “Tentative Standards for Prestressed Concrete Piles, Slabs, I-Beams and Box Beams for Bridges,” Joint AASHTOand PCI Committee Report, 1962.27 “Time Dependent Deflections of Pretensioned Beams,” Kelly, D.J. and others, CTR, Report 381-1, 1987.28 “A Study of Prestressed Panels and Composite Action in Concrete Bridges Made of Prestressed Beams,Prestressed Sub-Deck Panels and Cast-in-Place Decks,” Furr, H.L., and others, TTI, Reports 145-1, 2, 3, & 4F, 1970& 1972.29 “Optimum Design or Reinforcement for Notched Ends of Prestressed Concrete Girders,” Menon, G. and R.W.Furlong, CFHR, Final Report 196-1F, 1977.30 “Optimum Design or Reinforcement for Notched Ends of Prestressed Concrete Girders,” Menon, G. and R.W.Furlong, CFHR, Final Report 196-1F, 1977.31 “A Computer Program to Analyze Beam Columns Under Moveable Loads,” Matlock, H. and T.P., Taylor, CTR,Report 56-5, 1968.32 “A Rational Method for Estimating Camber and Deflection of Precast Prestresseed Members,” Martin, L.D., PCIJournal, Jan./Feb. 1977.33 “Time Dependent Deflections of Pretensioned Beams,” Kelly, D.J. and others, CTR, Report 381-1, 1987.34 “Design of Continuous Highway Bridges with Precast, Prestressed Concrete Girders,” C.L. Freyermuth, PCIJournal, April 1969.35 “Automated Design of Continuous Bridges with Precast Prestressed Concrete Beams,” Jones, H.L. and others,TTI, Volume II, Program Documentation, Report 22-1F, 1974.36 “Design of Simple-Span Precast Prestressed Bridge Girders Made Continuous,” NCHRP Project 12-29, FinalReport, Est. 1989.37 “Iron and Steel Beams,” 1883-1952, American Institute of Steel Construction, Fifth Printing 1968.38 “Ten-Division Influence Lines for Continuous Beams,” Eighth Edition, Dr. G.A. Ing, Frederick Zinger PublishingCo., New York, 1956.“Moment Shears and Reactions for Continuous Highway Bridges,” American Institute of Steel Construction, NewYork, 1966.39 “The Analysis of Continuous Beams for Highway Bridges IV,” Georgia D.O.T., 1971, Enhanced by SDHPT.40 “Guide Specifications for Horizontally Curved Highway Bridges,” AASHTO, 1980 with 1986 revisions.41 “Time Dependent Deflections of Pretensioned Beams,” Kelly, D.J. and others, CTR, Report 381-1, 1987.42 “Effects of Bracing on I-Girder Bridges,” Schelling, AISC Project 308.


Chapter 8 Substructure Design

Contents:Section 1 — Abutments ........................................................................................................8-3

Section 2 — Interior Bents....................................................................................................8-8

Section 3 — Piers................................................................................................................8-35

Section 4 — Pier Protection................................................................................................8-41

Chapter 8 — Substructure Design Section 1 — Abutments




Section 1 Abutments

Overview

Abutments present special problems in bridge design. They make design methodsproblematic, and they complicate detailing and construction. However, most bridges requirethem. Abutments must be compatible with the bridge approach roadway. They must havebackwalls to keep the embankment from covering up the beam ends and to support possibleapproach slabs. They usually have wing walls to keep the sideslopes away from thestructure and to transition between the guard rail and the bridge rail.

Background

Timber piling with timber lagging to hold back the embankment were once the cheapestsolution. A number of these still exist on county roads across the state. Many have lowstructural ratings because of bending in the abutment piling.

Early concrete abutments were called U type or cantilever type. U type abutments had twoside walls and a front wall resting on spread footings below natural ground. The side wallswere long enough to keep the embankment from encroaching on the bridge opening. Thetaller the abutment, the longer the sidewalls. Cantilever type abutments had variable widthrectangular columns supported on spread footings below natural ground. The fill was builtaround the columns and allowed to spill through, on a reasonable slope, into the bridgeopenings. A cap was placed on top of the columns to support the superstructure, andearwalls were added to the ends of the cap to keep the fill away from the bearing area. Agreat number of these types of abutments were constructed in Texas, and they performedvery well. Detailing and construction problems, however, were severe and so was the cost.

By 1940, most of the abutments in Texas were of the “stub” or “perched” type, constructedby driving piling or drilling shafts through the finished fill and placing a cap backwall andwing walls on top. The header bank was sloped from the top of the wing wall through theintersection of the cap and backwall into the bridge opening. The bridge must beconsiderably longer than with U type abutments but slightly shorter than with cantilevertypes. Figure 8-1 shows this relationship. The extra length was justified on the basis of costand aesthetics.



Figure 8-1. Abutments (Online users can click here to view this illustration in PDF.)



Although more economical, stub abutments are associated with certain maintenanceproblems. The well known “bump at the beginning of the bridge,” caused by fill settlement,is hard on abutments. Horizontal movement of the fill has caused torsion cracking in thecaps. Longer wing walls require a pile or drilled shaft to support their weight. If thesesupports are stopped too close to natural ground, the settling fill can drag the wing walldown, breaking it off at the cap.

The design of abutments with backwalls has been standardized through trial and error.Vertical moments and shears in the cap are insignificant because of the participation of therelatively deep backwall and because support spacing is kept smaller than for interior bentsto increase horizontal resistance. Abutments are generally the most complicated detailingproblem in a bridge.

There have been some attempts to secure the abutment to the superstructure, usually withunsatisfactory results.

Current Status

Most bridges designed in Texas have “stub” abutments. Some important features that maybe included in the design and construction of abutments include the following:

♦ Wing walls

♦ Retaining walls

♦ Approach slabs

Wing Walls. A wing wall confines the abutment backfill material and roadway soil at thesides, behind the abutment backwall. Wing walls can be cantilevered or founded. Thelength of the cantilevered wing wall is limited to 12 ft. Wing walls greater than 12 ft. inlength must be founded by drilled shaft(s) or pile(s). The available Texas Department ofTransportation (TxDOT) “Standard Details” for abutments include wing wall details. Thesedetails should be adhered to if the case is appropriate. If designing a wing wall, be sure toinclude a live load surcharge in the analysis. Reinforcing bars that tie the backwall and thewing wall together must extend on each side of the joint enough to develop the strength ofthe bar. Additional information can be found in the TxDOT Bridge Detailing Manual.

Retaining Walls. Because of increased urban rehabilitation work and restricted right-of-way, retaining walls are often required at the ends of bridges. In some cases these walls maypass by the sides of the abutment cap, eliminating the wing walls. In other situations, thewall may cross in front of the abutment cap. Commonly, walls pass alongside and in frontof the abutment creating a U-shaped wall.

Several types of walls may be used in conjunction with bridge abutments. In cut situations,the walls will often be cantilevered drilled shaft type walls, tied-back walls, or even spreadfooting type walls. The wall and bridge abutment will often become a single structure inthese cases. Design of this type of retaining wall abutment can become complex, and shouldbe a cooperative effort between bridge and geotechnical engineers. Soil or rock nailed wallsmay also be used to support abutments in cut situations. With nailed type walls, the



abutments are usually standard type “stub” abutments that are completely separate from thewalls.

In the most common, fill type situation, the walls will usually be mechanically stabilizedearth (MSE) walls. Although the abutment cap can be placed directly on the MSE fillwithout deep foundations, this has not been a common practice in Texas, therefore drilledshaft or piling foundations must be provided. The foundations are required to be installedprior to construction of the MSE wall, in order to avoid damage to the wall reinforcementsduring foundation installation.

Approach Slabs. An approach slab is a 13 in. thick lightly reinforced concrete slab thatprecedes the abutment at the beginning of the bridge, and follows the abutment at the end ofthe bridge. Its intended purpose is to provide a smooth transition from roadway pavement tobridge slab. The use of approach slabs is optional, and some districts have had success withtheir use, while others have had success without their use.

If used, the Bridge Design Section strongly discourages supporting the approach slab onwing walls. Experience has shown that compaction of the backfill is difficult, and that theloss of backfill material can take place. Without the bearing on the backfill, the approachslab becomes a slab supported on three sides, at the two wing walls and the abutmentbackwall. The standard approach slab is not reinforced for this situation, nor are the wingwalls designed to carry the load.

It is suggested that the approach slab should be supported by the abutment backwall and theapproach backfill only. Therefore, an appropriate backfill material is essential. TxDOT iscurrently supporting the placement of a cement stabilized sand (CSS) “wedge” in the zonebehind the abutment. CSS solves the problem of difficult compaction behind the abutment,and is resistant to the moisture gain and loss of material that is common under approachslabs. The use of CSS has become standard practice in several districts and has shown goodresults. Contact the Bridge Design Section or Geotechnical Branch for suggested limits forthe CSS.




As a general rule, the use of 40 pounds per cubic foot equivalent fluid pressure with 2 ft. ofsurcharge, if no approach slab is used, will suffice for calculating horizontal forces.Retaining type abutments in questionable soils may justify a more accurate analysis.

Recommended design practice for standard type “stub” abutments with backwalls is asfollows:

♦ The position of the backwall, wing wall lengths, wing wall support, and various otherstandardized items should be as shown in the TxDOT Bridge Detailing Manual or theapplicable bridge standard drawings. Cantilever wing walls should be used wherepossible.

♦ Cap, backwall, and wing wall reinforcing should also conform to the TxDOT BridgeDetailing Manual. Structural analysis will not be required.

♦ For pile foundations, battered pairs of piling should be used for all abutments that arenot otherwise restrained from horizontal movement. Examples of sufficient restraint areslab spans and pan form spans that are doweled into the abutment, and abutments withina mechanically stabilized fill. Battered piling should never be used adjacent to MSEwalls because of the difficulty of installing the backfill.

♦ The maximum spacing of drilled shafts or pile groups should not exceed 16 ft. withbeams 40 in. and less in depth nor 12.5 ft. with beams of greater depth.

♦ Drilled shaft loads may be calculated as the total vertical load on the cap dividedequally among the cap shafts. Wing wall shaft or pile load is usually taken as 10 tonsper shaft or pile.

♦ Pile loads may be calculated as the total vertical load on the cap divided equally amongthe cap piling plus the load caused by 40 pounds per cubic foot fluid pressure from thebottom of the cap to 2 ft. above the roadway surface. The back pile should not beallowed to go into tension due to the lateral load.

♦ A construction joint should be provided in abutment caps longer than 90 ft. The jointshould clear the bearing seat areas.

Chapter 8 — Substructure Design Section 2 — Interior Bents


Section 2 Interior Bents

Background

The design of interior bents is more direct than it is for abutments. Load paths are cleaner,and lateral earth pressure is not usually a factor. Wheel loads can be followed through thebeams into the cap and columns. Moments and shears are calculated at various locations forcomparison to strength of the member. The loads are carried on down to the foundation anddelivered to the earth.

In Texas, timber trestle bents were the earliest type, followed by concrete walls and thenconcrete columns with cap. Early engineers put their ideas of aesthetics into the columnsand caps.

Trestle pile bents, after the 1930s, had steel or concrete piling with concrete caps. Concretepiling were reinforced with mild steel until the 1950s when prestressed piling took over.Prestressed piling are easier to handle and more durable. Steel piling have been usedextensively but have been susceptible to severe corrosion at the ground line. Many havebeen fitted with concrete collars to correct the problem.

Multiple-column bents are composed of a concrete cap, two or more concrete columns, andhidden foundations. Until the late 1940s, most columns were square. They often had largechamfers or radii at the cap and occasionally had decorative collars or were tapered in width.Now, most columns are round. Metal forms are required to ensure a smooth and even finish.Column sizes are limited to 6 in. incremental diameters for standardization and reuse offorms. Spiral reinforcement is used around the main steel, although they are designed astied columns. Originally, spirals were #3 at 3 in. pitch. This was soon increased to 6 in.pitch because of concrete placement difficulties and later to #4 spirals at 9 in. pitch incolumns with longitudinal bars that are #11 and larger. Square columns are still usedoccasionally in situations where aesthetics are considered. Engineers have chosen round,square, rectangular, rounded end rectangular, and, occasionally, other shapes in attempts toenhance the appearance of the bridge. Special surface finishes such as exposed aggregate,form liners, or texture paint have also been used.

Bent caps have been, and still are, basically rectangular reinforced concrete beams. In theearly days they tended more to chamfers, radii, and other decorative features. Now they aremostly prismatic except that the cantilever soffits are sloped. Single-column bent capsusually have sloping soffits. Inverted tee caps were introduced in the 1970s to reduce theclutter of deep caps in congested environments. They have been used in multiple- andsingle-column bents. Some have a sloping soffit but many are constant depth. Structuralsteel box girders have been used for caps on bents that require a long span between columnsto clear a roadway below. These caps are considered fracture critical, which requires carefulattention to details. No failure has been experienced. Post-tensioned concrete caps invarious forms have also been used in these situations.

Typical trestle pile and multiple column bents are shown in Figure 8-2.



Single-column bents came into use on ramps and connectors in urban interchanges. Theyhelp alleviate the congested appearance by minimizing the number of columns, althoughsome of the effect is lost because single columns are larger than multiples. Variations inshape have been more extensive for single columns. Rectangular is the basic shape withcircular or chamfered end faces as variations. They may be prismatic or tapered in onedirection. Hollow sections have been used for extremely large columns.

Design of reinforced concrete columns is very complex. Column members are subject tosimultaneous axial compression and bending. Accurate determination of the amount ofbending is especially difficult. When a column deflects due to any primary cause, asecondary moment is generated in the amount of the axial load times the deflection, whichin turn generates a little more deflection. An iteration process is required to reach acondition of stability. Deflections are inversely proportional to moment of inertia andmodulus of elasticity, neither of which is subject to accurate determination in reinforcedconcrete members, which may be cracked to varying degrees at different loading stages anddifferent locations along the column length. Manageable analysis procedures have beendeveloped based on magnification of the moments due to external loads using simplifiedassumptions for elastic behavior of the column. These uncertainties have apparentlyresulted in conservative designs since no column malfunctions in Texas can be attributed tothe effects of service loads.

Design of caps is more straightforward, usually as a reinforced concrete beam.

Pile footings are designed as beams subject to considerations in the “Foundations” section ofthe American Association of State Highway and Transportation Officials (AASHTO)Specification.

Construction Issues

Construction problems have been caused by congestion of reinforcing. Too many tie bars toallow efficient vibration of the concrete was a common complaint at one time. Lap splicesin main reinforcing can impede concrete flow. Internal stiffening is often required forstability of the reinforcing cage during erection and concrete placement, which canaggravate already crowded conditions.



Maintenance Issues

Maintenance problems have primarily been caused by catastrophic or corrosive conditions.Columns up to 30 in. in diameter can be demolished by an errant truck. If there are only twocolumns per bent under simple spans with open joints, the bridge will fall. On one occasiona 54 in. column was sheared by an airborne tank truck from a ramp above. The reinforcingsteel was offset but unbroken, and the column continued to support the bridge. Columnshave also been weakened by scour and moved by drift or migrating soils. Columns underleaking joints in bridges on which salt is used for ice removal will, in time, lose theirconcrete cover layer due to corrosion of the reinforcing steel. Fire has also caused loss ofconcrete cover due to expansion of the steel and, occasionally, internal heat damage to theconcrete.

Caps designed by load factor methods can have very high dead load steel stresses. This hascaused alarming cracks in a few caps that required epoxy injection for appearance andcorrosion protection.



Figure 8-2. Typical Interior Bents (Online users can click here to view this illustration inPDF.)



Current Status

Trestle pile bents are acceptable for stream crossing structures. Design heights are usuallylimited by lateral soil properties and the bending strength of piling regularly used.

Multiple-column bents are used in the majority for both stream crossing and gradeseparations. Caps may be rectangular or inverted tee with or without sloping cantileversoffit. Columns are usually round but may occasionally be square or rectangular. Columnsize may change within the bent height, producing a multi-tiered bent. A tie beam may beprovided at the size change if necessary to control sidesway moments and unbraced columnlength. Multi-tiered bents with web walls are considered to be piers. See Section 3, Piers, inthis chapter for information on piers.

Single-column bents are recommended for interchanges and elevated highways whereaesthetics dictate a minimum of columns. Inverted tee caps are often used for prestressedbeam spans. Rectangular caps are recommended with continuous steel girder units, ifclearance permits. If clearance to the cap is critical, a steel cap framed into the girders canbe used. A concrete cap cast around and prestressed to the steel girders has been used. Thepreferred column shape is rectangular and prismatic. It is conceded that column shape willbe a controversial item on most interchange projects. Except for historical and decorativebridges, acceptable aesthetics can be achieved by eliminating as many columns as practicaland maintaining smooth lines with unobtrusive caps.

Post-tensioned concrete or steel box beam caps are recommended for straddle bent capsbeyond the span limit of reinforced concrete.



Design Recommendations – Rectangular Caps

Trestle pile and multiple-column caps may be analyzed as beams on knife edge supports atthe center of piling or columns. In the middle 1960s a bent cap analysis program (CAP 18)was developed at the University of Texas on a research project.1 The program is usedregularly for multiple-column bents and can be used for single-column bents. Assumptionsrecommended for use with this program or with longhand analysis are as follows:

♦ Dead load reactions due to slab and beam weight are applied as point loads at centerlineof beam. This applies to prestressed and steel, box, and I- and U-shaped girders. Inshort, cantilever regions of the cap actual bearing pad locations for box and U-beamsmay need to be considered. Dead loads due to railing, sidewalks, medians, and overlaycan usually be distributed evenly to all the beams.

♦ Live load plus impact reactions per lane are based on the governing truck or laneloading. For lane loading the 18 kip concentrated load should be used. The use of thelighter 18 kip force for interior bent reactions has been the philosophy of the departmentsince the 1950s. Note that the 26 kip force is used when the uniform load extends inone direction only (e.g., abutment foundation load calculations). Typically the totalreaction is then modeled as shown in Figure 8-3. The live load is distributed to thestringers assuming the slab hinged at each stringer, except the outside stringer. Laneboundaries should be carefully considered so as to produce maximum stress at variouscritical locations along the cap.

Figure 8-3. Model of Total Live Load Reaction per Lane for Bent Cap Design (Online userscan click here to view this illustration in PDF.)

♦ Class C concrete (f ′c = 3,600 psi) and Grade 60 reinforcing steel are typically used.Higher concrete strengths are sometimes used in large caps supporting very long spans.

♦ Caps are typically 3 in. wider then the columns. Cap depth should be in 3 in.increments and not less than the cap width.



♦ A construction joint should be used in multicolumn bents when the distance betweenoutside columns exceeds 80 ft. The joint should be located close to a dead loadinflection point but not under a bearing seat buildup.

♦ Flexure design: use the strength design method (load factor) with the check fordistribution of flexural steel using z = 170 for moderate exposure conditions. The useof 130 for z is typically only justified for bridges over coastal waterways. In addition,it is our policy to limit the stress in the steel due to unfactored dead loads to 22 ksi tofurther minimize cracking that has been observed in numerous bent caps in Texas.This 22 ksi limit will usually control the final design on longer span structures.

♦ Design negative moments are taken at the effective face of the column. Location ofeffective face requires engineering judgment but is generally the face of a square orrectangle, based on equivalent area, for round or irregular shaped columns.

♦ Maximum and minimum reinforcement ratios should be limited by the requirements inAASHTO.

♦ While #9 or #10 bars are sometimes used, in the majority of cases #11 bars are used tosimplify design and construction. Mixing of bar sizes is usually not justified.

♦ Typically the minimum number of bars is four top and bottom, and the maximumnumber in a layer is limited by a 2 1/2 in. clear spacing requirement to facilitateconcrete placement and vibration. A second layer may be placed 4 in. on center fromthe outside layer. A third layer should only be used in very deep caps. A horizontal tiebar tied to the vertical stirrup legs should support second and/or third layers. In heavilyreinforced caps, bundling bars in two-bar bundles is sometimes used to maintainnecessary clear spacing. Layered and/or bundling bars should comply with AASHTO.

♦ For most caps anything in excess of four top bars can be cut off in compression zonesbetween columns. To simplify design, bars should usually be conservatively extendedLd past an inflection point rather than adhering to the complex requirements inAASHTO. For bottom reinforcement, limit the number of bars across a column andinto a cantilever to three or four to avoid congestion with vertical column steel.Additional bars should end at the column face. These top and bottom bar cut-offcriteria apply to conventional caps with moderate amounts of reinforcement. For largecaps with heavy reinforcement, follow the provisions in AASHTO.

♦ Bars longer than 60 ft. will require laps. Attempts should be made to locate these lapsin compression or very low tension zones. Lap lengths should be based on tension laprequirements (see TxDOT Bridge Detailing Manual). Consideration should be given tostaggering or alternating laps in adjacent bars to minimize congestion. Mechanicalcouplers or welded splices may be specified for stage construction. For additionalinformation concerning reinforcing steel, or stage construction, refer to the following:• Chapter 5, Section 1: Materials• Chapter 4, Section 5: Stage Construction



♦ Many cantilevers are too short to allow full development length for the #11 Grade 60top reinforcement. However, research conducted at the University of Texas in the mid1960s indicated that the reaction from the outside beam provides a clamping effect andthat a bar extension of 15 in. beyond the center of the beam will develop the bar. Thestandard distance from centerline beam to end of cap is 1 ft.-9 in., which should beconsidered as a minimum for new designs.

♦ Longitudinal skin reinforcement should be used in accordance with AASHTO in capsdeeper than 3 ft. Caps 3 ft. and less should have two #5 bars equally spaced in eachside face.

♦ Shear design: use the strength design method (load factor) with Vc = 120bd.

♦ For most conventional caps, use #5 stirrups with a 3 in. minimum and 12 in. maximumspacing. Double stirrups may be required close to column faces. For large heavilyreinforced caps, #6 stirrups are typically used.

♦ Between columns, design for shear at the face of the column and extend required stirrupspacing to a convenient distance beyond the centerline of the beam but not less than1 ft. This is slightly more conservative than AASHTO Specifications but simplifiesdesign. Try to minimize the number of stirrup spacing changes.

♦ In cantilever regions, shear need not be considered unless the distance from center ofload to effective face of column exceeds 1.2d. Provide stirrups at 6 in. spacing. Forlonger cantilevers with sloping soffits, d may be taken at the inside edge of bearing.

A detailing item that merits attention from the design engineer is the bearing seat build-upfor prestressed beam spans. Extreme grades and skews can produce conflicts between thebearing seat or bent cap and the beams or bearings if the seats are not shown properly on thebent details. The TxDOT Bridge Detailing Manual shows typical bearing seatconfigurations. Note that bearing seat build-ups taller than 3 in. require reinforcement,which should be shown on the detail.

Design Recommendations – Inverted Tee Caps

Inverted tee caps have been the subject of considerable research.2 Appendix B illustrates therecommended design method. The design recommendations for rectangular caps shownabove also apply to inverted tee caps. Other salient features of inverted tee cap design are asfollows:

♦ Since the caps are usually deeper than 3 ft., beam side reinforcing should be providedaccording to Figure 8-4.

♦ Primary moment and shear design is similar to that for rectangular caps. For moment, bis the bottom width for negative moment and top width for positive. For shear, b is thestem width.

♦ Distribution of reinforcement and dead load stress limits are the same as for rectangularcaps.

♦ Since the bearings are relatively far from the center of the cap, torsion should beconsidered in single-column bents, but it will rarely control.



♦ Ledge depth and reinforcing is determined by punching shear, shear friction, ormoment.

♦ Web reinforcing must be sized for hanger loads and for vertical shear, possibly incombination with torsion. Hanger load stresses, which usually control, are not added toshear and torsion stresses. Extra vertical reinforcing should be provided across theend surfaces of the stem to resist cracking, which has been observed in existingbridges.

♦ The former practice of welding bars together for development of ledge reinforcing hasbeen discontinued. It has been determined that widening the ledge and reinforcingsimilar to Figure 8.4 provides sufficient development.

Figure 8-4. Beam Side Reinforcement, Inverted Tee Caps (Online users can click 8-4 to viewthis illustration in PDF.)



Design Recommendations – Steel Box Girders

Steel box girders may be used if the cap design span is beyond the practical limits ofreinforced concrete. The usual occurrence is on two-column bents straddling a lowerroadway. Box girders may be below the span beams if clearance permits, but the usual casehas simple span beams supported by brackets on the sides of the box girders, so the exposeddepth can be minimized. Box girder caps have been pierced by continuous steel I-girders,but this practice is discouraged because it can create a terribly fatigue-prone detail. Boltedconnections must be used for this condition. Since single- or two-column steel bent caps areconsidered fracture critical, bad fatigue details must be avoided. Figure 8-5 shows a provendetail for attachment of prestressed beam support brackets. Box girders may be painted orbe of weathering steel. Painting inside the box girder is controversial. For several years,recommendations of the Bridge Design Section were to weld the girders air tight and notpaint inside for either regular or weathering steel. Current practice is to provide a removablehatch cover at both ends, for access by fracture critical inspectors, and to paint the insideregardless of whether the outside is painted or bare. Anchor bolts can be located inside thebox to improve aesthetics. Three options for web-to-flange welding have evolved toaccommodate different fabricators’ procedures (see Figure 8-5 and Figure 8-6).

Items to consider are as follows:


♦ Web plates should preferably be designed to resist shear and bending without transversestiffeners. In extreme cases it may be desirable to add shears due to St. Venant torsionaccording to Prestressed Concrete Structures3 or Theory of Simple Structures.4

♦ Overturning should be considered. The contractor should be advised if certain erectionor slabbing sequences will not be permitted.

♦ Deck slabs continuous across steel box girder caps are discouraged.

♦ Drip beads and pans may be required to protect concrete from weathering steel run-off.



Figure 8-5. Steel Box Girder Bent Cap Example, Cap Supporting Prestressed Beams(Online users can click here to view this illustration in PDF.)



Figure 8-6. Steel Box Girder Bent Cap Example, Access Hole (One End Only) (Online userscan click here to view this illustration in PDF.)



Design Recommendations – Integral Steel I-Girders

Interchange structures often contain continuous steel girder units requiring a minimum ofcap projection below the bent cap for vertical clearance reasons. Because of horizontalclearances, single-column bents are also usually required. The integral or framed steel bentcap has been used many times in many ways in the past. I-girders are the overwhelmingchoice of shape because of difficult framing details and questionable fatigue performance ofpierced box girders. Prior to the early 1980s, steel girders were framed into steel bent capsusing field welded connections. This is virtually impossible to do without creating aCategory E detail in the girder. Although there have been no catastrophic failures in 25years of usage, numerous cracks have been found and it is currently considered moreappropriate to use bolted connections. Figure 8-7 shows an acceptable detail for boltedintegral steel bent caps. If weathering steel is used for steel girder units, the same isrecommended for integral bent caps. Items to consider in design are as follows:


♦ Web plates preferably should be designed to resist shear and bending without transversestiffeners.

♦ Bearing stiffeners should be used over floorbeam reactions.

♦ Provision for rotation due to girder deflection should be provided, but the bearingshould be fixed against translation.

♦ Transverse overturning should be carefully considered. Any fluctuation of tension innon-prestressed anchor bolts must be within the allowable stress for a Category E detail,regardless of the anchor bolt strength.

Design Recommendations – Integral Prestressed Concrete

Projects have been built in Texas with concrete caps surrounding and post-tensioned tocontinuous steel I-girders. This design is supposed to be more consistent in appearance withapproaching inverted tee concrete caps and avoid the complicated framing details of a steelcap. There is a disadvantage in erection of the steel girders because the interior columnscannot be used for support until the cap is stressed. It remains to be seen if there aresufficient advantages to make this solution desirable. At this time, the Bridge DesignSection does not recommend this method.



Figure 8-7. Girder Connections, Integral Steel Girder Bent Caps (Online users can clickhere to view this illustration in PDF.)



General Design Recommendations for Columns

Accurate column analysis is virtually impossible. Vertical loads are statistical, horizontalloads are estimates, and the manner in which the forces due to these loads distribute throughthe structure is indeterminate. Column response to these forces is influenced by incipientbuckling and secondary bending, which depend on flexural section properties usuallyvarying with the forces. Much has been written and researched on columns. Design ofSlender Non-Prismatic and Hollow Concrete Bridge Piers5 contains a good discourse oncolumn design. Specifications have adopted conservative approximations of column actionfor use by the designer. Additionally, computer programs6 such as BMCOL 51 cancalculate secondary effects directly based on input section properties. The computerprogram PIER7 is the most advanced reinforced concrete column analysis tool available tolocal designers. It uses a fiber model to analyze any shape, solid or hollow, using input non-linear properties of the concrete. Secondary effects are calculated and biaxial bending istreated accurately.

Vertical loads are reactions from the bent cap design loads. Impact is included. Horizontalloads are wind, longitudinal, and centrifugal forces as described earlier (see Chapter 6,Section 2). For single-column bents and single-tier, multiple-column bents, longhandcalculations will usually suffice for moments due to horizontal forces. Columns on singledrilled shafts may be assumed fixed at three shaft diameters (but no more than 10 ft.) belowthe top of the shaft. For single-column bents on footings with two drilled shafts, transversefixity is assumed at the top of footing and longitudinal fixity at three shaft diameters (but nomore than 10 ft.) below the bottom of the footing. Analytical studies have shown that fixityis usually achieved with considerably less embedment, even in soft soils. However, someconsideration should be given to the possibility of some future minor excavation around theshafts. For bents in erodeable stream beds, scour has to be considered. Some of thecomputer programs that attempt to predict scour are overly conservative, and consultationwith a qualified geotechnical and/or hydraulic engineer is recommended to give a betterestimate as to the top of the ground for column design purposes. In any case, if the designerwishes to make a more accurate prediction of the depth of fixity, lateral load-deflectionproperties of the soil can be input into a nonlinear laterally loaded foundation program suchas COM624. Columns on footings with multiple drilled shafts or piling in both transverseand longitudinal directions are considered fixed at the top of footing.

Moments may be magnified for design according to the specifications or computed withprograms that account for P∆ effects.

Vertical loads and moments are factored and the section strength compared by an ultimatestrength computer program8 PCACOL or charts shown in Appendix C. Biaxial bending isconsidered in terms of a load contour method (not AASHTO) unless the program PIER isused.

Strong consideration should be given to designing columns with Grade 40 reinforcing steeleven though 60 ksi steel is now required for all reinforcement on TxDOT projects.Experience indicates that the design of the majority of bridge columns is controlled by theAASHTO minimum reinforcement ratio of one percent. Stress requirements can usually besatisfied with a Grade 40 design and significantly shorter embedment, and splice lengths can



then be justified. This is primarily a constructibility issue and is the philosophy used in thecurrent state standards.

For columns on a single drilled shaft, the critical section will usually be in the drilled shaft.If the section is adequate for 3,000 psi concrete, Class A concrete may be allowed in thedrilled shaft. If 3,600 psi is required, Class C concrete should be specified for shafts by plannote.

Designers are encouraged to take comfort in past performance and avoid complicatedstudies in column design whenever possible. It is evident that considerable redundanciesexist, even for single-column bents, which make it difficult for bridge columns to bend farenough to break. Recognition of conditions that deserve extensive analysis is characteristicof experienced designers.

Earthquake effects are not considered at all in the design of columns by the Bridge DesignSection.

Column Design – Multi-Column Bents

Most of the columns under Texas bridges have 3,600 psi concrete reinforced with Grade 40reinforcing steel. Occasionally, higher strength concrete is required.

Round columns are currently used for a majority of structures. They are greatly preferredfor multi-column bents with rectangular caps. Available sizes, minimum and usualreinforcing, and recommended height limits are shown on Figure 8-9. Within these heightlimits, and with reasonable column spacing as used on the state standards, except for veryunusual conditions, the following column sizes may be used without analysis for axial loadand bending.

Slab spans 24 inchPan form spans 24 inchPrestressed beam spans

Types A, B, C beams 30 inchTypes IV, VI(Mod) beams 36 inch

When analysis for axial load and bending is necessary, horizontal forces must be resolvedinto components parallel and perpendicular to the bent by longhand, since there is noreliable computer program available to the Bridge Design Section. Design of single-tierbent columns can be completed by modeling individual columns as fixed against rotationand deflection at some assumed fixity point (see discussion on fixity underGeneral Design Recommendations for Columns) into the soil while free to rotate and deflectin the longitudinal direction. In the transverse direction the top of the column should beassumed to be free to translate but not rotate. Moments can be magnified to account forslenderness (P∆) effects by using the manual method described in the AASHTOSpecifications. However, this method is highly conservative and should only be used when



a computer program, such as BMCOL 51, that considers secondary effects is not available.Factored loads and moments may be compared to column strengths obtained from theinteraction curves of Appendix C.

Square columns are occasionally used for aesthetic enhancement of a structure. Experiencehas been insufficient to justify size selection without calculation. Design methods may bethe same as for round columns except that transverse and longitudinal moments must bemagnified separately and combined in an interaction equation. Effective length factors, k,may be taken as 1.0 transversely and 1.5 longitudinally. Factored loads and moments shouldbe compared to column strength using the computer program PCACOL. Alternatively,secondary effects and biaxial bending may be investigated using the PIER program.Recommended tie reinforcing is shown in Figure 8-9.

Round and square columns in multi-tier bents should be analyzed with the FRAME 11program (sidesway allowed). Transverse and longitudinal moments should be magnifiedseparately. Effective length factor may be taken as 1.0 transversely. Longitudinally, use ofthe BMCOL 51 program is encouraged because of the interaction of different lengths andcolumn sizes in different tiers. Desirable column to tie beam connection details are shownin Figure 8-10.

Multiple-column bent tiers with web walls could be considered braced in the transversedirection, but this is immaterial since there will be no transverse moment to magnify in thebraced tier. Column capacity in the longitudinal direction is not considered affected by theweb wall.



Figure 8-8. Minimum Column Steel Requirements for Round Columns (Online users canclick here to view this illustration in PDF.)



Figure 8-9. Examples of Typical Square Columns (Online users can click here to view thisillustration in PDF.)



Figure 8-10. Column to Tie Beam Connections, Multi-Tier Bents (Online users can clickhere to view this illustration in PDF.)



Column Design – Single-Column Bents

Single columns are usually rectangular with square, circular, or chamfered ends. They maybe tapered in the transverse or longitudinal width but not in both. Extremely large columnsare often hollow. For column heights over 100 ft., consideration should be given to windloads more appropriate for the location and height. Wind Forces on Structures9 may be ofassistance in this regard. Longhand methods are appropriate for application of horizontalloads and calculation of column loads and moments in single-column bents. Longitudinaland transverse moments must be magnified separately using the AASHTO Specificationmethods or BMCOL 51. Effective length factor may be taken as 2.0 in both directionsunless it can be shown, analytically, that restraints provided by the superstructuresufficiently limit secondary moments. Column strength may be estimated using PCACOL.Alternatively, for prismatic solid columns and preferably for non-prismatic or hollowcolumns, the PIER program can be used to consider secondary effects and biaxial bending.Typical single columns are shown in Figure 8-11. A once-used hollow column section isshown in Figure 8-12.

Steel Columns

Virtually the only steel columns likely to be used on Texas bridges are H-shaped trestlepiling. The height capabilities of these are fairly well established by past practice usingstandard details. Should a more accurate analysis be required, loads and moments may becalculated in a manner similar to that described above for multiple-column concrete bents.Stresses due to the unfactored loads and moments should be compared to the service loadallowables given in the AASHTO Specification.

Prestressed Concrete Columns

Besides prestressed concrete trestle piling, there have been other columns prestressed forcrack control or to allow precast segmental construction. It has occasionally appeareddesirable to extend trestle piling to greater than standard heights. For these situations, it isrecommended that loads and moments be calculated in a manner similar to reinforcedconcrete columns above. Unfactored and unmagnified loads and moments should produceno more tensile stress than allowed by the AASHTO Specification. Factored and magnifiedloads and moments should not exceed the ultimate capacity of the cracked section. The PCIDesign Handbook10 contains ultimate strength interaction curves for square members up to24 in. with eight seven-wire strands. Beyond this size and when mild steel reinforcing isused in addition to prestressed reinforcing, longhand methods can be used assuming 0.003concrete strain and linear variation of strain with steel stresses limited to 60 ksi for mildreinforcing and the average stress in prestressing steel at ultimate load.



Figure 8-11. Single Column Examples (Online users can click here to view this illustrationin PDF.)



Figure 8-12. Hollow Column Example (Online users can click here to view this illustrationin PDF.)



Foundation Loads

Foundation loads must be based on service load design.

It is a policy of the Bridge Design Section to calculate foundation loads based on theaverage vertical load on each column. Thus, for typical multiple-column bents the verticalload at the bottom of each column is the weight of the bent plus dead load reaction of thesuperstructure plus the live load reaction, without impact, of the maximum number of lanesthat can occupy the roadway width times the lane reduction factor, all divided by the numberof columns.

The rationale for this is that slight settlement of the foundation will cause redistribution ofactual column loads, tending toward equalization. Also, if different loads are given for eachcolumn in a bent, there is a probability that foundation tip elevations will be set differentlyfor each column. This practice is strongly discouraged because of differential settlementpossibilities.

For single drilled shaft foundations, the average load at the bottom of a column is the“calculated drilled shaft load,” which will be shown on the plans and used in the foundationdesign.

For individual pile footings of typical multiple-column bents, the footing weight is added tothe average bottom of column load and is divided by the number of piling in the footing.Theoretical pile load variation due to moment at the top of the footing is ignored except forunusual situations. This is given in tons as the “calculated pile load” for the bent. Use ofthis method will result in the same number of piling in each footing of a bent.

Single-column bents usually have multiple shaft or pile footings, and the foundation loadsshould be the maximum, considering all appropriate loading groups and including momenteffects at the top of the footing. Extremely large or complicated footings may justify moreaccurate procedures, but usually it will suffice to calculate the loads in kips as follows:

Foundation Load =

PN

MxCxIx

MyCyIy+ +

P = Vertical load at bottom of footing (k)

N = Number of shafts or piling

M = Moment at top of footing (k ft.)

C = Centroid to extreme shaft or pile (ft.)

I = Moment of inertia of pile group about centroid (ft.2)

The maximum of the calculated loads divided by the allowable overstress for the causativegroup loading, converted to tons, is the “calculated drilled shaft load” or “calculated pileload” shown for the bent.



The type of foundation, and to some extent the allowable load, depends on the soil profile.A geotechnical engineer should be consulted before finalizing the foundation design.Maximum calculated loads recommended by the Bridge Design Section are given in the "Maximum Allowable Loads on Foundation Elements" table below. Refer to the TxDOTGeotechnical Manual for a more complete discussion.

Table 8.1: Maximum Allowable Loads on Foundation Elements♦♦♦♦ Maximum Allowable Load (tons)

Type Size Trestle Pile Pile FootingSteel H Piling 12 x 53 40 70

14 x 73 60 100

Concrete Piling 16 in. sq. 75 12518 in. sq. 90 17520 in. sq 110 22524 in. sq. 140 300

Drilled Shafts 30 in. dia. 27536 in. dia. 40042 in. dia. 52548 in. dia. 70054 in. dia. 90060 in. dia. 110066 in. dia. 130072 in. dia. 1500

♦ Based on service load designAbility of the pile or drilled shaft to transfer load to soil may limit these loads.

Note: Trestle piles should be checked for structural adequacy.

Footings

Service load design without impact is recommended for drilled shaft or pile footings. It willusually be acceptable to calculate moments and shears assuming all shafts or piling toexperience the maximum load simultaneously. Moments and shears are calculatedaccording to the rules given in the AASHTO Specification. Shear seldom controls except onextremely large footings. If the shearing stress should exceed the specification maximum, itmay be corrected by thickening the footing, adding shear reinforcement, or both.

Certain design concepts and load recommendations for top and side reinforcing, spacing,clearances, and embedment are illustrated in Figure 8-13 and Figure 8-14.



Figure 8-13. Typical Pile Footing (Online users can click here to view this illustration inPDF.)



Figure 8-14. Figure 8.14: Typical Drilled Shaft Footing (Online users can click here to viewthis illustration in PDF.)

Chapter 8 — Substructure Design Section 3 — Piers


Section 3 Piers

Background

Early piers were solid walls or webbed square columns, usually supported by two large-diameter caissons excavated to the founding elevation. Most of the caissons were excavatedby clamshell working inside. The weight and the cutting edge on the bottom allowed thecaisson to sink as the material inside was removed. A few were deep enough and wetenough to require excavation within a pneumatic chamber. The piers were sturdily designedto resist the destructive forces associated with rivers and bays. Piling replaced caissons, andconstruction was performed in a cofferdam usually formed with steel sheet piling. Usually,excavation was performed inside the cofferdam and piling were driven below the waterusing a follow block. Seal concrete was then placed around the piling. After curing of theconcrete, the caisson could be dewatered and the footing and shaft constructed in the dry.

Large-diameter drilled shafts in water became practical in the 1960s. Many river piers wereconstructed with a single shaft under each column. The shaft could be constructed from atemporary soil island or by using double casing and a floating drilling rig. The tops ofdrilled shafts were located about a foot above water or, if the pier was outside of the normalwater line, could be stopped a few feet below natural ground. A tie beam between shaftswas sometimes used, and columns were constructed on top of the shafts or tie beam. Webwalls were often constructed between columns to strengthen the pier against drift carried bya flood.

For piers adjacent to the Intracoastal Canal, the U.S. Coast Guard required a flush face onthe traffic side. These piers had square columns with a wall between.

After the Sunshine Skyway disaster in Florida, considerable attention was directed toprotection of piers from the consequence of impact from an ocean-going vessel.11 Texashas three bridges that were subject to these considerations. One has large dolphins to wardoff ship impact, and the others have large sand and rock islands built around the criticalpiers.

Current Status

Piers are considered to be any substructure composed of reinforced concrete walls orcolumns with full or partial height walls adjacent to a waterway. Large hollow columnsunder major waterway bridges are also called piers. On cable-stayed bridges, tower supportsand all other substructure under the unit may be referred to as piers. Most piers on majorriver crossings are constructed with round columns and web walls extending full height, asshown in Figure 8-15. These are called dumbbell piers. Piers adjacent to the IntracoastalCanal may be required to have a smooth face, as shown in Figure 8-16. Piers adjacent toocean-going traffic require special design against the catastrophic effects of ship impact.



Figure 8-15. Figure 8.15: Typical Dumbbell Pier (Online users can click here to view thisillustration in PDF.)



Figure 8-16. Intracoastal Type Pier (Online users can click here to view this illustration inPDF.)




A few design considerations will be given for dumbbell piers, intracoastal piers, and ocean-going channel piers followed by a discussion of seal design, which is common to all three.

Piers in rivers or other significant flowing waterways deserve careful attention byengineering personnel in the district, hydraulics group, geotechnical group, and bridgedesign group. Stability of the riverbanks and possible future channel migration should beconsidered. Future scour depth, including scour due to the presence of the pier, should beestimated by the best available methods and taken into account in the structural design.Orientation parallel to the stream flow is critical for minimizing scour and overturning forceof floodwater on the pier. Foundation elements must extend sufficiently below the scourline to resist design loads. At the anticipated scour depth, piers should have a minimumfactor of safety of 1.0 under Group II loading.

Dumbbell Piers. If dumbbell piers are used, web walls should extend above the elevation ofany possibility of drift in the area. Pier caps usually cantilever past the outside columns forappearance and consistency with interior bents in the bridge. The cantilever is designed asfor any multi-column bent, but the portion between columns can be nominally reinforced ifsupported by the web wall. Web walls should be no less than 1.25 ft. thick to facilitateconcrete placement. Web wall design is nominal and according to past practice. Typicaldetails are shown in Figure 8-15. Web walls are assumed to fully support connected capsand tie beams and to prevent transverse moment in the connected columns. They are notconsidered to affect the longitudinal section properties of the columns. Because these piersare more sensitive to longitudinal moments and more likely to be used with longer spans, itappears appropriate to use 19 pounds per square foot for longitudinal superstructure windinstead of the usual approximation.

Intracoastal Piers. Intracoastal Canal piers should be subject to the same designconsiderations given above for dumbbell piers. Columns are usually square so the crashwall can be connected smoothly. Crash wall design is nominal. Acceptable details areshown in Figure 8-16.

Ocean-Going Channel Piers. Piers adjacent to ocean-going channels are usually subject tomodified design procedures. Spans are long and high and usually constructed by unusualmethods, such as cantilever segmental concrete box girder or cable-stayed segmental,creating additional loads to the piers. Wind loads may be increased because of the heightand further modified for the shape of the superstructure based on wind tunnel tests.Magnitude and application of wind loads during cantilever erection conditions will benegotiated. Wind Forces on Structures.12 can be used as a guide Design controls for arecent cable-stayed bridge are shown in Figure 8-17. Ship impact should preferably beresisted by protective dolphins or islands. Piers should be sturdy near the water line, butdesigning for unimpeded impact from ocean-going vessels is considered impractical. Piersections will probably be single- or multi-cell concrete boxes, cast-in-place or precast andpost-tensioned. The span-to-thickness ratio for hollow columns is subject to controversy.The Bridge Design Section recommends a maximum ratio of 7.5 based on the researchreported in Design of Slender Non-Prismatic and Hollow Concrete Bridge Piers.13 Thisstandard has been relaxed when it was proven that the column capacity was more than twicethe factored load and moment.



Seal Design

Seal design should be based on hydrostatic pressure created by normal water elevation. Forriver crossings this level may be difficult to establish. The assumed elevation for designmust be shown on the plans. Arguments have occurred in the past when pier footings wereconstructed during high water and the contractor claimed TxDOT was liable for not havingprovided an adequate seal. The construction specification is now very specific in thisregard.

Seal concrete is assumed to weigh 150 pounds per cubic foot, and buoyancy is taken as 63pounds per cubic foot times the depth of water to the bottom of the seal. Piling or drilledshafts may be assumed to resist uplift in the amount of 10 pounds per square inch of contractarea. Friction between the seal and cofferdam should not be considered. Reinforcing steelis not used in seals. For extreme spacing of piling or drilled shafts, shear and bendingshould be investigated in the plain concrete section. Recommended allowable stresses are300 pounds per square inch in tension and 80 pounds per square inch in shear. Seals arealways placed with Class E concrete, which has six sacks of cement per cubic yard andminimum compression strength of 3,000 pounds per square inch.



Figure 8-17. Figure 8.17: Example Design Wind Loads for a Recent Long-Span Bridge(Online users can click here to view this illustration in PDF.)

Chapter 8 — Substructure Design Section 4 — Pier Protection


Section 4 Pier Protection

In coastal areas along ship channels, the Intracoastal Canal, navigable rivers, and bargecanals, piers need protection from marine traffic. Barge tows can generate large forces onimpact, but the energy released by a large ocean-going vessel in collision can beastronomical. Besides bracing the piers with crash walls or mass concrete as appropriate,local practice suggests fender systems for barge traffic and large dolphins or, preferably,sand and rock islands for ocean-going traffic.

Each situation will require careful consideration of the size and speed of marine traffic andoperating characteristics of the channel. The U.S. Corps of Engineers can furnish valuableinformation about this. Criteria for the Design of Bridge Piers with Respect to VesselCollision in Louisiana Waterways14 is a good general discourse on problems andprobabilities in the Gulf Coast area.

Fender systems have been used for many years. Some districts prefer timber and othersprefer steel, usually based on maintenance experience. Steel rusts badly, but some believethe average damage due to impact is less severe and easier to repair. Special coatings havebeen developed to protect steel in marine environments. Because of fit-up between thewales and driven piling, connections will usually be welded. This requires an elaboratecoating repair procedure that may still be the weak spot in the corrosion protection system.Casual impact will be a regular occurrence, as evidence suggests that barge operators like touse fenders as navigational aids under certain conditions. Currently, the U.S. Coast Guardrequires that non-sparking material be used for the horizontal wales that will contact thevessel to minimize the possibility of ignition of flammable material. Steel members musthave a timber or plastic facing. It has been determined that fender systems will sustain lessdamage if they consist of vertical piling only, instead of bracing with battered pilingaccording to past practice. Large timber pile clusters or small concrete dolphins are oftenused at the ends and angle breaks of the fenders. Large elastomeric energy absorbers areavailable to ease the force on the support members, but usage has been minimal in Texas.Given the forces to be resisted, accurate analysis of a fender system is very complicated.Computerization of Fendering Systems 15 gives an insight into this . General requirementsfor recent Intracoastal Canal fender systems are shown in Figure 8-18.

Design of pier protection against ocean-going vessels is even less defined. Much has beenresearched and written,16 but application to a particular situation is tedious. Goodinformation about the size, speed, and operational characteristics of marine traffic isessential. A probabilistic risk analysis is usually performed to determine the most desirablelevel of protection. Protection for the pier may be large dolphins or islands.17 Protection formotorists may be required in the form of warning signs, lights, horns, or barricades actuatedmanually or by span failure detectors.

Federal Highway Administration (FHWA)-coordinated research,18 ( using pooled fundsfrom several coastal states, attempted to develop an acceptable guide specification for pierprotection.

Design of any pier protection system should be referred to the Bridge Design Section.



Figure 8-18. Typical Fender Configuration (Online users can click here to view thisillustration in PDF.)



1 “A Computer Program to Analyze Bending of Bent Caps,” Matlock, H. and W.B. Ingram, CFHR, Research Report56-2, 1966.2 “Design of Reinforced and Prestressed Concrete Inverted T-Beams for Bridge Structures,” Furlong and Mirza, PCIJournal, July – August 1985.3 “Prestressed Concrete Structures,” Line, T.Y. and Ned H. Burns, Third Edition, John Wiley and Sons, Inc.4 “Theory of Simple Structures,” Shedd, T.C. and J. Vawter, John Wiley and Sons, Second Edition, 1966.5 “Design of Slender Non-Prismatic and Hollow Concrete Bridge Piers,” Poston, R.W., and others, CTR, FinalResearch Report 254F, 1983.6 “A Computer Program to Analyze Bending of Bent Caps,” Matlock, H. and W.B. Ingram, CFHR, Research Report56-2, 1966.7 “Design of Slender Non-Prismatic and Hollow Concrete Bridge Piers,” Poston, R.W., and others, CTR, FinalResearch Report 254F, 1983.8 “Strength Design of Reinforced Concrete Sections,” Portland Cement Association, Computer Program.9 “Wind Forces on Structures,” J.M. Briggs, Chmn., Transactions, American Society of Civil Engineers, Paper No.3269, Volume 126, Part 2, Final Report, 1961.10 “PCI Design Handbook,” Third Edition, Prestressed Concrete Institute, 1985.11 “Pier Protection for the Sunshine Skyway Bridge,” Knott and Flanagan, IABSE Colloquium, Preliminary Report,1983.12 “Wind Forces on Structures,” J.M. Briggs, Chmn., Transactions, American Society of Civil Engineers, Paper No.3269, Volume 126, Part 2, Final Report, 1961.13 “Design of Slender Non-Prismatic and Hollow Concrete Bridge Piers,” Poston, R.W., and others, CTR, FinalResearch Report 254F, 1983.14 “Criteria for the Design of Bridge Piers with Respect to Vessel Collision in Louisiana Waterways,” Modjeski andMasters, Report to LADOTD and FHWA, 1985.15 “Computerization of Fendering Systems,” Derucher, Heins and Schelling, American Society of Civil Engineers,“Computing in Civil Engineering,” Conference Paper, June 1978.16 “Criteria for the Design of Bridge Piers with Respect to Vessel Collision in Louisiana Waterways,” Modjeski andMasters, Report to LADOTD and FHWA, 1985.“Bridge and Pier Protection Systems,” Derucher and Heins, Marcel Derucher, New York 1979.“Ship Collisions with Bridges,” Committee on Ship-Bridge Collisions, Marine Board, National Research Council,1983.17 “Bridge and Pier Protection Systems,” Derucher and Heins, Marcel Derucher, New York 1979.18 “Vessel Passage at Navigable Waterway Bridge Crossings,” Knott, M.A. and others, Draft Report, Task A, forFHWA, 1988.


Chapter 9 Special Designs

Contents:Section 1 — Widenings ........................................................................................................9-2

Section 2 — Strengthenings................................................................................................9-13

Section 3 — Deck Replacements........................................................................................9-17

Section 4 — Raisings..........................................................................................................9-19

Section 5 — Railroad Underpasses.....................................................................................9-29

Section 6 — Pedestrian Underpasses..................................................................................9-49

Section 7 — Historic Bridges .............................................................................................9-53

Section 8 — Long Span Bridges.........................................................................................9-54

Section 9 — Bridge Railing ................................................................................................9-61

Section 10 — Expansion Joints ..........................................................................................9-62

Section 11 — Bearings .......................................................................................................9-72

Section 12 — Anchor Bolts ................................................................................................9-89

Section 13 — Deck Drainage..............................................................................................9-97

Section 14 — Reinforced Concrete Box Culverts ............................................................9-101

Section 15 — Reinforced Concrete Pipe ..........................................................................9-105

Section 16 — Corrugated Metal Pipe ...............................................................................9-109

Section 17 — Structural Plate Structures..........................................................................9-111

Section 18 — Long Span Structural Plate Structures .......................................................9-112

Section 19 — Sign Support Structures .............................................................................9-113

Section 20 — High Mast Illumination Poles ....................................................................9-117

Section 21 — Traffic Signal Poles....................................................................................9-119

Section 22 — Sound Walls ...............................................................................................9-121

Section 23 — Wildlife Issues............................................................................................9-122

Chapter 9 — Special Designs Section 1 — Widenings


Section 1 Widenings

Background

Widening of bridges began in the 1950s and has been a steady source of work for designersand contractors. In 1963, widenings accounted for about 3 percent of the total deck area ofbridges let to contract. In 1988, this number was 10 percent. In 1997, bridge wideningaccounted for about 5 percent of the total bridge construction cost.

The various types of bridges have been widened as follows:

♦ Old slab spans were widened with slab spans.

♦ Old concrete girder spans were widened occasionally with slab spans but usually withsmaller concrete girders or with prestressed concrete beams. Pan form girders,prestressed box beams, and double tee beams have been used in isolated instances.

♦ Pan form girders and prestressed beam spans are usually widened in kind.

♦ Steel I-beam spans were widened with steel until it was decided that prestressedconcrete beams could be used without the difference in stiffness causing deck distress.Prestressed beams have since been used extensively for widening steel beam spans andeven cantilever/drop-in and continuous steel units.

♦ Steel I-beam spans with timber piling have been widened and the exposed portions ofthe timber piling removed, under traffic, and replaced with concrete posts or steel Hpiling supported by a footing cast around the timber piling below ground.

♦ Continuous steel I-beam units have been widened in kind but more often withprestressed beams continuous for live load or with a continuous deck on simple-spanprestressed beams.

♦ Cantilever/drop-in steel I-beams have been widened with simple-span prestressedconcrete beams. The deck expansion joint is over the interior bents in the widenedportion with longitudinal open joints connecting to the existing deck joint at thecantilever end.

♦ A continuous steel plate girder unit (100-140-100 ft.) was widened with cantilever/drop-in prestressed beams with the deck continuously reinforced across the notched ends.

♦ Trusses have not been widened, except for one continuous deck truss on which therequired width was attained with cantilever frames from each panel supportinglongitudinal stringers.

♦ Farm-to-market road slab spans provided a widening challenge, as discussed in SimpleSlab Spans in Chapter 7, Section 3 of this manual. There is a way to widen the H10high curb design to make it theoretically adequate for H20 loading, but the method is socomplicated as to make it impractical.

Because of bridge deck deterioration, many widenings currently include replacement of theexisting slab.



Substructure design has taken many forms, some of which create a strange appearance. Capshave been extended with different size caps. Caps have not been extended and the new beamsupported on a column only. Round columns have been used to widen square columns oreven trestle pile bents. Round columns with web wall have been used to widen solid wallpiers and tapered square column/web wall piers. The possibilities appear to be endless.

Spread footings have been used for widening foundations when the existing bridge is onspread footings. Trestle piling, pile footings, and drilled shafts are also used. Drilled shaftspredominate.

Design can be interesting because of the many variations in geometry and structure. Untilrecently, most bridge widening plans were made by district personnel from design sketchesprepared by the Bridge Design Section. Now, most plans are prepared in the Bridge DesignSection and the metropolitan districts.

Construction Issues

Construction problems have occurred because of differential dead load deflection betweenthe new beam and the existing. If diaphragms are in place between the new and old beams,they may be broken loose. There seems to be no appreciable effect on the slab even whenplaced very close to heavy traffic. Limber columns with no tie to the existing cap haveallowed uncomfortable lateral deflection before the deck connection is made. Span lengthvariation in old bridges has caused the designed structure to misfit. Breakback of existingconcrete can cause damage to the remaining structure if done improperly. Deck slabs notbroken back over a beam are particularly susceptible to damage underneath.

Maintenance Issues

Maintenance problems are rare. Most widened deck slabs have an asphaltic concrete overlaythat tends to hide any surface cracking. Deck problems have not been reported. There havebeen a few cases where the new foundation, being drastically different from the old, causeddifferential vertical movement. The extreme case involved existing spread footings in aswelling clay with well anchored drilled shafts in the widening. Upward movement of thespread footings, acting through the new diaphragms, raised the new beams clear of thebearing seat.

In another bridge the spread footing settled due to drying of the founding material, loadingup the pile foundation in the widening and causing extensive cracking.

Widening one side only, or otherwise unsymmetrical with the existing centerline, createsvariable and thick overlay to accommodate the new crown. Thick overlay is susceptible torutting.

Current Status

Bridge widening can be an economical improvement to roadway traffic and safetyconditions. A load rating and bridge condition survey must be completed before plans are



begun. The survey must evaluate the condition of all structural elements, especially thebridge deck. Chloride concentration must be measured if there is evidence that the deck hasbeen salted. Farm-to-market slab (FS) spans of any loading should not be widened.


Design guidelines for the various elements of new bridges may also be applied to widenings.Additional observations are given below.

It is generally conceded that existing structures with an Inventory Rating of H20 and aboveare suitable for widening. An Inventory Rating of H17 is sometimes acceptable forwidening structures with low average daily traffic (ADT). Ratings should be based on thecurrent American Association of State Highway and Transportation Officials (AASHTO)Specification except that allowable stresses should be based on the minimum materialstrengths used on the original construction. Allowable stresses for structural steel may betaken from the "Chronology of Simple Steel I-Beam Standards" table or for standard designsfrom the "Chronology of AASHTO Specification Requirements for Structural Steel PlateGirders Era" table . Section properties of old I-beams are given in Iron and Steel Beams.1 Allowable stresses for reinforcing steel may be taken from the "Chronology of ReinforcingSteel Specifications (1918-1953)" table and the Chronology of Reinforcing SteelSpecifications (1957-1988 Interim) table. Allowable concrete stresses may be as shown forslabs on the "Chronology of AASHTO Specification Requirements for Concrete SlabsReinforced Perpendicular to Traffic" table.

Widened portions should be designed for HS20 loading.

♦ Slab spans can be widened with slab spans: Skewed slabs with main reinforcingperpendicular to the bents will be weak if the edge beam is removed under traffic. Theedge should be shored under this condition. Alternatively, dowels can be grouted intothe existing slab edge and the widening placed with reinforcing parallel to the centerlineof roadway. Curbs may be removed after the new slab has cured.

♦ Concrete girder spans can be widened with concrete girders, prestressed beams,prestressed box beams, or double tees: Prestressed beams are recommended. Boxbeams may be used if depth is a problem.

♦ Pan form girders can be widened with pan forms, prestressed box beams, or doubletees: Pan forms are recommended. Alternates using double tees may be appropriate forcertain situations.

♦ Steel I-beam spans can be widened with prestressed beams or steel I-beams:Prestressed beams are recommended. Steel I-beams may be used if depth, framing, oraesthetics is a problem.

♦ Continuous steel I-beam units can be widened with prestressed beams or steel I-beams:Simple-span prestressed beams with the slab continuous are recommended. The slabshould have standard reinforcing and be tied to the existing slab.

♦ Cantilever/drop-in steel I-beam units can be widened with prestressed beams orcontinuous steel I-beams: Simple-span prestressed beams are recommended with



expansion joints over the bents connected by longitudinal open joints to the existingexpansion joint at the notches.

♦ Continuous steel plate girder units can be widened with continuous steel plate girdersor with prestressed beams if the span is 140 ft. or less.

♦ Prestressed concrete beam spans and units should be widened in kind.

Some examples of typical widening details are shown in:

♦ Figure 9-1

♦ Figure 9-2

♦ Figure 9-3

♦ Figure 9-4

♦ Figure 9-5

♦ Figure 9-6

♦ Figure 9-7



Figure 9-1. : No Longer Recommended Practice of Widening of FS Concrete Slab Spans(Online users can click here to view this illustration in PDF.)



Figure 9-2: Examples of Widening Concrete Girder Spans (Online users can click here toview this illustration in PDF.)



Figure 9-3. Figure 9.3: Examples of Widening Pan Form Girder Spans (Online users canclick here to view this illustration in PDF.)



Figure 9-4: Examples of Widening Using Double Tee Beams (Online users can click here toview this illustration in PDF.)



Figure 9-5.: Examples of Widening Steel I-Beam Spans (Online users can click here to viewthis illustration in PDF.)



Figure 9-6.: Example of Widening Cantilever/Drop-In Steel I-Beam Unit with PrestressedConcrete Beams (Online users can click here to view this illustration in PDF.)



Figure 9-7.: Example of Widening Continuous Steel I-Beam Unit with Prestressed ConcreteBeams (See following explanatory notes. Online users can click here to view this illustrationin PDF.)


Longitudinal deck reinforcing is continuous across interior supports. Interior diaphragms arenot required. End diaphragms are required at ends of units only.

Chapter 9 — Special Designs Section 2 — Strengthenings


Section 2 Strengthenings

Background

Strengthening has not been a prevalent practice for Texas highway bridges.

Early widenings of H15 I-beam bridges often required cover plates or angles welded to thebottom flanges. Under the growing emphasis of fatigue design, it became apparent that thesestrengthening details could create other problems that are worse than theoretical overstress.Current fatigue specifications make it virtually impossible to weld cover plates to old or newsteel beams.

Dire need to retain portions of existing structures for traffic handling during construction hasresulted in retrofitting some cover plate ends by bolting as shown in Figure 9-8. Usually,there have been other bad fatigue details in the structure that make it desirable to measureactual stress ranges in the bridge before strengthening.

Many steel girder bridges with noncomposite decks have been redecked with new compositeslabs with the addition of shear connectors.

Steel trusses have been strengthened to accommodate heavier decks. Bolting was used toattach plates or shapes to improve moment capacity of chords carrying direct deck load or todecrease slenderness of compression members. Rusted rivets were replaced with high-strength bolts.

Concrete superstructures are hardly ever strengthened. Concrete columns have beenreinforced by encasement. This is usually a maintenance procedure required by damage dueto chloride attack, fire, or vehicular impact.

There have been other maintenance procedures to restore strength lost to design error,material deficiency, construction mistakes, and hostile environment. Such procedures arealways done under pressure, challenge ingenuity, and are expensive.

A few examples are given below:

♦ Design failure to evaluate erection stresses due to curvature of continuous steel girders:Required additional shoring during erection to remain in place until the concrete slabhad cured.

♦ Design error in sizing bent cap shear reinforcing: Required external hanger boltinstallation.

♦ Lower concrete strength in prestressed beams than required by the design: Beamswere shored near mid-span until the concrete slab had cured.

♦ Structural steel girder flange plate found to contain extensive laminations and non-metallic inclusions: Long cover plate was bolted to the flange after erection of thegirder.



♦ Bridge deck slabs constructed with insufficient effective depth: Concrete overlay wasbonded to deck.

♦ Foundation elements mislocated during construction: Various solutions includedstraddle footings, revised cap design, and strengthening of footings.

♦ Pin and hanger damage due to wear and skew on redundant structures: Replace pinand hangers under traffic.

♦ Suspected pin fracture on non-redundant structure: Install temporary hangers andreplace pins while closed to traffic.

♦ Brittle fracture in a redundant steel girder unit: Install temporary hangers and reweldthe flange under partial traffic.

♦ Out-of-plane bending cracks in a skewed steel girder unit: Gouge and weld cracks, andweld stiffeners to the flange in the vicinity of bents.

♦ Steel beams hit by overheight load: Heat straighten, replace sections of web and flange,or replace sections of girder depending on severity of damage.

♦ Prestressed concrete beams hit by overheight load: Dry pack, pneumatically grout,epoxy inject, or replace beam depending on severity of damage.

♦ Columns broken by vehicular impact: Shore and encase or replace column and part ofcap depending on severity of damage.

♦ Columns damaged by fire or chloride intrusion: Remove soft concrete. Epoxy groutsmall dowels and encase. If spiral is corroded, install new ties before encasement.

♦ Bents undermined by scour. Piers moving with the river bank: Construct outboardfoundation and transfer superstructure reaction to new cap. Insulate new columns frommoving river bank with compressible materials.

♦ Steel piling corroded at ground line. Concrete piling broken or exploded: Encase withconcrete.

♦ General softening of concrete piling due to sulphate attack: Construct new outboardfoundation to pick up the load on prestressed footing beams.



Figure 9-8. : Example of Retrofit Cover Plate Ends (Online users can click here to view thisillustration in PDF.)

Current Status

Strengthening to increase load-carrying capacity should be carefully evaluated in terms offatigue susceptibility and cost. Repairs should be coordinated with the Bridge DesignSection, Bridge Construction Branch, or bridge engineers in the metropolitan districts.




Generally, structural members with an inventory rating of H17 or better may remain in placewithout strengthening. Rating should be based on AASHTO Specification service loadmethods. This is considered justified by the conviction that a more complicated and accuratestructural analysis, such as with a computerized grid program, would reveal an excess ofstrength.

New composite decks are often added to steel girder bridges. Steel beams and girders withCategory D, E or E' fatigue-prone details should also be evaluated for “remaining usefullife” before significant modifications are made. Equipment exists in the Bridge DesignSection for field measurement of stress range histograms and prediction of remaining life.

Chapter 9 — Special Designs Section 3 — Deck Replacements


Section 3 Deck Replacements

Background

Since the 1960s, when deck deterioration was recognized as a serious problem, there havebeen an increasing number of bridge decks replaced.

Most deck replacements are accompanied by bridge widening. Generally, if the chloridecontent in the second inch below the top of the old slab was 2 pounds or more per cubic yardof concrete, the existing deck was replaced. If the chloride content in the second inch wasless than 2 but the top inch was 2 or more, it was considered sufficient to scrabble off aninch and overlay the existing deck with 2 in. of concrete. In isolated cases, prestressed beamstructures with sufficient structural capacity have been overlaid with a full depth reinforcedconcrete slab in lieu of replacement. Chloride-laden concrete was removed to a maximumdepth of 6 in. in one continuous slab bridge, replaced with Class C concrete, and overlaidwith 2 in. of dense concrete. Shoring was required because of exposed negative momentreinforcing. Deck replacement on concrete girder spans has not been considered practical.

When decks are replaced on non-composite steel I-beam or plate girder spans, shearconnectors are usually installed on all existing beams.

Construction of 2 in. overlays is tedious, especially with the dense concrete type, whichrequires special mixing and screeding equipment. The continuous slab that required shoringalso was widened on the outside and in the median and was on a horizontal curve, all ofwhich created a unique and complicated structure to build. Removal of decks appears to beno great problem. Advancements in field stud welding methods have eased the addition ofshear connectors.

Maintenance procedures have not yet been required, although some of the earlier 2 in.overlays have some cracking and possible delamination.

Current Status

Decisions to replace or rehabilitate a deteriorated bridge deck should be based on the resultsof a load rating and a thorough condition survey. A complete discussion of bridge loadratings and condition surveys is presented in Chapter 3 of the Bridge Project DevelopmentManual. Decks on older noncomposite steel girder bridges are often replaced, regardless oftheir condition, as a means of strengthening the girders. The new, thicker decks are madecomposite with the girders through the addition of stud connectors.


New deck slabs shall be designed as provided in Chapter 7 of this manual. See Chapter 7,Section 25 for design recommendations for steel I-beam span shear connectors, and seeChapter 7, Section 26 for design recommendations for steel plate girders.

Chapter 9 — Special Designs Section 3 — Deck Replacements


Although seldom used, general warrants for overlaying an existing deck with a full-depthreinforced slab are:

♦ The existing structure should be composite.

♦ Concrete over the beam should be sufficiently sound to provide some composite actionfor dead load. The width and thickness to be used in the analysis will be a matter ofjudgment based on the deck condition.

♦ The beam should not be overstressed according to current allowables, considering aportion of the old slab composite for new slab load plus the usual width of new slabcomposite for live load.

♦ The old deck should be scrabbled before placement of the new slab. Dowels betweenold and new are not considered necessary to develop composite action.

Chapter 9 — Special Designs Section 4 — Raisings


Section 4 Raisings

Background

In the early days of freeway and interstate highway construction, vertical clearance of 14 ft.-6 in. was allowed. Although the maximum legal height of highway vehicles in Texas is13 ft.-6 in. (14 ft. for automobile carriers), permitting of overheight loads is routine anddesirable to facilitate the movement of equipment within the state. Many high loads aremoved without a permit, and others are higher than that stated on the permit. For variousreasons the low clearance structures on certain routes have demonstrated a propensity forhigh load impact damage. Sometimes the perpetrator is apprehended and required to makerestitution. Even so, there is usually significant disruption of highway personnel andhighway users until the damage is repaired.

Required vertical clearance was established at 16 ft.-6 in. for interstate highways in 1957 toprovide for national defense movements. It was soon adopted for all major highways. Thisalleviated the damage problem but did not eliminate it.

Many highway bridges have been raised, mostly over interstate routes. Vulnerability todamage from overheight loads is often the motivation. General rehabilitation of a section ofhighway may also require raising the crossovers to full 16 ft.-6 in. clearance above afreeway that may have been raised by pavement strengthening or gradeline improvement.

Lifting points must be realistically analyzed for highway bridges. Continuous units raised adifferent amount at each support require more complicated analyses and usually acceptanceof some theoretical overstress. Pedestal design has been somewhat variable.

Construction problems usually involve raising the bridge. Jacking from falsework adjacentto traffic is dangerous. Simultaneous raising of continuous beams requires coordination ofmany jacks.

Maintenance requirements have not appeared. None of the structures raised has been hitagain. Although some of the pedestal supports lack the sturdy appearance of the formerbearings, to date none have collapsed.

Current Status

Clearance-deficient prestressed concrete beam and steel beam or girder bridges are raised asdeemed appropriate by the responsible highway engineers.


Raising of bridge types other than prestressed beam or steel is likely to be as expensive asreplacement and is generally not recommended.



Steel and prestressed beam bridge raising can be cost effective. Skew and horizontalcurvature can be accommodated.

Jacking from falsework is discouraged for safety reasons. Jacking from bent caps is usuallypractical. Liberal analyses of prestressed beam diaphragms will usually reveal them capableof carrying jack reactions to the beams. This has been proven for most diaphragm details byfield experience. An exception is the thickened slab end allowed by current prestressedconcrete beam span standards. Experience has shown that these can be separated from thebeam during lifting. Two jacks per beam space are recommended. Steel bridges are easilyfitted with welded jacking supports, if the diaphragms are not suitable.

Continuous steel units can accommodate some variation in the amount adjacent supports areraised. Load factor design could be invoked to establish the allowable variation.Strengthening of the main members is not recommended because of the possibility of doingmore harm than good.

Steel pedestals are recommended between the cap and bearing for both steel and prestressedconcrete beams. Compressive stress in the pedestal should not exceed the specificationallowable for concentrically loaded columns. The resultant of dead load reaction andhorizontal force applied at the top of the pedestal should fall within the middle third of thebase plate width to prevent overturning.

Various types of steel pedestals have been used. The most common type is fabricated fromrolled W or HP shapes to minimize welding. One such design used on a recent continuoussteel I-beam raising is shown in Figure 9-9, Figure 9-10, and Figure 9-11. These probably donot conform to the above design recommendations for overturning. The steel I-beam unit,eight beams wide, was raised, then widened with four prestressed concrete beams on eachside and a new deck slab placed over all 16 beams. The prestressed beam bearings wereassumed to transfer most of the horizontal load. Figure 9-12 and Figure 9-13 show pedestalsfor continuous steel I-beam units on another project. The raised bridge was five beams wideand not widened. These figures indicate types that can be adapted to different designconditions.

Pedestals for prestressed concrete beams are shown in Figure 9-14, Figure 9-15, andFigure 9-16. Bracing between adjacent pedestals has often been used for lateral stability fortaller pedestals, particularly those exceeding 2 ft.-6 in.

Currently, square or rectangular steel tubing is recommended for pedestal design, because asmoother appearance can be achieved. See Figure 9-16.



Figure 9-9. : Example Pedestal Details for Raised Steel Beams – End Expansion (Onlineusers can click here to view this illustration in PDF.)



Figure 9-10. : Example Pedestal Details for Raised Steel Beams – Interior Expansion(Online users can click here to view this illustration in PDF.)



Figure 9-11. : Example Pedestal Details for Raised Steel Beams – Interior Fixed (Onlineusers can click here to view this illustration in PDF.)



Figure 9-12. : Pedestal Details for Raised Steel Beams – Example 2, View 1 (Online userscan click here to view this illustration in PDF.)



Figure 9-13. : Pedestal Details for Raised Steel Beams – Example 2, View 2 (Online userscan click here to view this illustration in PDF.)



Figure 9-14. : Example Pedestal Details for Raised Prestressed Concrete Beams – View 1(Online users can click here to view this illustration in PDF.)

Chapter 9 — Special Designs Section 5 — Railroad Underpasses


Section 5 Railroad Underpasses

Background

A railroad underpass allows vehicular traffic to pass under railroad traffic. A basicillustration of a railroad underpass and an example layout sheet can be found in Chapter 2 ofthe Bridge Detailing Manual.

In 1905, the American Railway Engineering and Maintenance of Way Association publishedthe Manual of Recommended Practice for Railway Engineering and Maintenance of Way asguidance to aid individual railroad companies develop their own policies and practices forrailroad design. In 1911 the association changed its name to the American RailwayEngineering Association (AREA) and published several subsequent manuals.

The association has since changed its name back to American Railway Engineering andMaintenance of Way Association (AREMA) and currently publishes the Manual forRailway Engineering (1999), with the similar goal of providing a guide from which railroadcompanies can develop their own policies and practices for railroad design. The AREMAManual for Railway Engineering governs railroad bridge design, in a similar fashion to theAASHTO Specifications for highway bridges.

Most railroads were in existence before the highway system was created. During the earlydevelopment of the highway system, railroad personnel or consultants performed mostrailroad underpass designs over state highways. These early designs usually consisted of ameager opening framed by steel beam spans and vertical abutment walls.

During the 1950s the task of designing railroad underpasses over state highways began totransition from the railroad company to the Texas Department of Transportation (TxDOT).

During the past 35 years, which is as far back as the records can be documented, there havebeen approximately 150 railroad underpasses let to contract over state highways. All ofthese designs were performed by TxDOT or a consultant retained by TxDOT.

Superstructure and substructure elements of railroad underpasses have been composed ofvarious materials and have included many different structure types. Some examples of thematerial and structure types that have been used are as follows:

Deck Types. Types of deck have included the following:

♦ Open timber

♦ Closed timber

♦ Steel plate

♦ Cast-in-place reinforced concrete

Beam Types. Types of beams have included the following:



♦ Steel I-beam (deck)

♦ Steel I-girders (deck and through)

♦ Precast prestressed concrete beams (deck) types A(Mod), C, C(Mod), IV, 34 Spl, 44Spl, 54, 54(Mod), 60, and 72

♦ TxDOT precast prestressed concrete box beams (deck) in depths of 34, 46, and 52 in.

♦ AREMA 42 in. reinforced concrete box beams (deck)

♦ Cast-in-place reinforced concrete slab

♦ Cast-in-place post-tensioned concrete gull wing girder

♦ Prestressed concrete gull wing girder

♦ Precast prestressed box beams (through)

Interior Support Types. Types of interior supports have included the following:

♦ Solid wall piers

♦ Closely spaced round columns with cap

Abutment Types. Types of abutments have included the following:

♦ Stub type with rip-rap protected slopes

♦ Retaining wall type, either cantilevered drilled shaft or cantilevered spread footing

Note: Square-up approach slabs are used for both abutment types on skewed structures.

Foundation Types. Types of foundation have included the following:

♦ Spread footing

♦ Trestle pile

♦ Foundation pile

♦ Drilled shaft

Recent Use. The most economical railroad (RR) bridges are constructed with prestressedconcrete I-beams. Prestressed concrete I-beams were used only on about 30 percent of theabove underpasses because of:

♦ The relatively short span for a given depth that prestressed concrete beam bridges cancarry or

♦ The railroad companies’ concerns over difficulties in repairing damaged prestressedconcrete beam bridges quickly with minimal disruptions in rail service

Through steel plate girders were disallowed for a while because of fear of failure caused bya derailed train. Recently, however, through steel plate girder bridges have been used for alarge percentage of RR bridges because they have been the best way to carry long spanswithin the vertical clearance constraints.



Examples of typical sections for some of the railroad underpass types that have beenconstructed in Texas are illustrated in the following figures:

Prestressed concrete beams Figure 9-17Through steel I-girders Figure 9-18Gull wing girders Figure 9-19TxDOT box beams Figure 9-20 and

Figure 9-21TxDOT and AREMA box beams Figure 9-22A unique system Figure 9-23A system with common details Figure 9-24

The letting year and the low bid unit cost for the entire underpass are included in the abovefigures. The costs may not be completely reliable due to economic conditions, biddingabnormalities, and locations of the structures. However, the figures should produce a senseof the range of structural possibilities, likely variables, and generally relative economy.



Figure 9-17. : Prestressed Concrete Beam Railroad Underpass (Online users can click hereto view this illustration in PDF.)



Figure 9-18. : Through Steel I-Girder Railroad Underpass (Online users can click here toview this illustration in PDF.)



Figure 9-19. : Gull Wing Girder Railroad Underpass (Online users can click here to viewthis illustration in PDF.)



Figure 9-20. : TxDOT Box Beam Railroad Underpass – Example 1 (Online users can clickhere to view this illustration in PDF.)



Figure 9-21. : TxDOT Box Beam Railroad Underpass – Example 2 (Online users can clickhere to view this illustration in PDF.)



Figure 9-22. : TxDOT Box Beam/AREMA Box Beam Railroad Underpass (Online users canclick here to view this illustration in PDF.)



Figure 9-23. : Railroad Underpass – Unique System (Online users can click here to viewthis illustration in PDF.)



Figure 9-24. : Railroad Underpass – System with Common Details (Online users can clickhere to view this illustration in PDF.)

Current Status

TxDOT or a consultant retained by TxDOT now designs virtually all railroad underpassesover state highways. Railroad companies are reimbursed for the cost of checking thesedesigns.

The TxDOT Bridge Division-Bridge Design Section, and the TxDOT bridge projectmanager, should be closely involved with any discussions with the various railroadcompanies concerning railroad bridges designed and built with federal and state funding foror by TxDOT.



Railroad right-of-way, and all structures within the railroad right-of-way, are the property ofthe individual railroad companies. This means that anything done on or across railroadright-of-way is subject to their review and approval. A formal railroad agreement must beexecuted prior to any such work. The Bridge Project Development Manual containsinformation concerning project lead times and agreement requirements.

Complete agreement between the railroad company and the TxDOT Bridge Division shouldbe obtained regarding the design features before beginning any design work. Review by therailroads of the design and plan details for railroad work should be coordinated through theTxDOT bridge project manager.

Superstructure Type

TxDOT prefers the following superstructure types, in order of preference:

1. Prestressed concrete I-beams with concrete deck (30 degree maximum skew),approximate 2000 cost = $125/sf

2. Box beams with concrete deck (15 degree maximum skew), approximate 2000 cost =$150/sf

3. Steel I-beams or steel plate girders with concrete deck on top (30 degree maximumskew), approximate 2000 cost = $175 to $200/sf

4. Steel through girders with steel floor beams and concrete deck (30 degree maximumskew), approximate 2000 cost = $225 to $275/sf

Rolled-in superstructures (complete superstructures placed on the substructure as a unit)have generally not been allowed because the railroads were concerned that TxDOT’scontractors could not complete them in a short enough period of time (usually 3 to 4 days).Work is done during short interruptions of rail traffic, and any delays in opening a structureto rail traffic can be very costly to a railroad.

Concrete box culverts, usually precast boxes that are jacked and bored under the track so asnot to disrupt rail traffic, are another commonly acceptable railroad supporting structure.

Other structure types not mentioned above may be acceptable to the railroad company butmust be negotiated on a case-by-case basis with the TxDOT Bridge Division and therailroad company involved. It should be noted that the cast-in-place post-tensioned trough,which was pioneered by Southern Pacific Railroad, is no longer allowed by Union PacificRailroad because of the concern that it would be difficult to repair and the repair woulddisrupt train movements.

Railroad traffic handling is an important part of the overall design that is usually carefullynegotiated during preliminary planning for the underpass. Information concerning thehandling of rail traffic can be found in Chapter 4 of the Bridge Project DevelopmentManual.

The TxDOT Bridge Division has the in-house expertise to turnkey design railroad bridgesand all appurtenances, or assist in the negotiations with a railroad company for the design



features of any railroad bridge or associated items. This wealth of knowledge andexperience can help assure the successful outcome of any railroad project.

Recent Examples

Features and typical sections of three recent railroad underpasses designed by TxDOT areincluded below:

1. A steel deck girder bridge designed in 1999 for the Union Pacific Railroad is shown inFigure 9-25. Note the 1 ft.-6 in. clear distance between bottom beam flanges and theslab extending to the bottom of the top flange, both requirements unique to UnionPacific Railroad. The ballast curb is a modified T-501 traffic rail shape and the roadbedis wide enough to accommodate the required walkways for trainmen. This simplifiessuperstructure details over the cantilevered walkways shown in Figure 9-24, but itbrings the live load closer to the wing walls, thus possibly resulting in a heavier wingwall design.

2. A steel through girder bridge designed in 1996 is shown in Figure 9-26. This type ofstructure is used where the apparent superstructure depth is required to be held to aminimum for spans over approximately 40 ft. The clearance width was increased inaccordance with AREMA requirements for a structure on a curve. The structuredepicted in this figure meets all Union Pacific Railroad requirements except those for abolted built-up bottom flange. On structures with low vertical clearance, Union PacificRailroad has required in some cases that there be no welded connections to the bottomor tension flange because of repairability and fatigue considerations. Fatigue concernsare for notches in the tension flange from over-height load impacts.

3. Rail traffic must run throughout the duration of a bridge replacement project.Sometimes this necessitates the construction of a temporary, or shoo-fly structure on analternate alignment. A shoo-fly bridge superstructure is shown in Figure 9-27. Theseare usually open deck structures, with the ties connected directly to the beams withhook bolts. Beams are usually short span wide flanges with no paint and the ties,walkways, and handrails are untreated lumber. Substructures are Class A concrete withno surface finish, and the beams bear on the substructure with only a masonry plate (nobearing pad or shoe). The general notes must be written to waive these requirements inthe TxDOT Standard Specifications. These structures are temporary, so they aredesigned to minimize the expense to construct and facilitate dismantling of the structureupon project completion. Since they are temporary they may be designed for Cooper E-72, with railroad approval, and AREMA fatigue requirements are usually waived.Fatigue-prone details should still be avoided, however.



Figure 9-25. : Example Steel Deck Girder Bridge (Online users can click here to view thisillustration in PDF.)



Figure 9-26. : Example Steel Through Girder Bridge (Online users can click here to viewthis illustration in PDF.)



Figure 9-27. : Example Typical Section of a Shoo-Fly Railroad Underpass (Online userscan click here to view this illustration in PDF.)


The American Railway Engineering and Maintenance of Way Association’s Manual forRailway Engineering governs railroad bridge design and shall govern all railroad structuredesigns, unless a railroad company has design guidelines other than, or in addition to, theAREMA guidelines.



Union Pacific Railroad has a written set of design requirements that must be adhered inaddition to the AREMA Manual for Railway Engineering. The Union Pacific Railroaddesign requirements must be maintained on all Union Pacific Railroad work unless variancecan be secured before design begins. No other railroad company currently has a writtenbridge design policy, but they all have their own requirements that must be followed. UnionPacific Railroad design requirements will generally suffice for most railroads.

Since a railroad company’s design policy and practice may differ from those presented inthe AREMA Manual for Railway Engineering, all design requirements must be determinedby the bridge designer during the preliminary planning stages of the project since theseadditional requirements may greatly affect the design.

The Texas Standard Specifications (1993) govern the construction of railroad bridges andare generally approved as satisfactory to meet the requirements of most railroad companies.The one exception is in the area of structural steel fabrication, in which case thespecifications refer to the AREMA provisions, Chapter 15, Section 3. A copy of TexasStandard Specifications (1993) should be included along with the design notes and plandetails for railroad review. Additionally, structural steel fabricators must be made aware ofthe AREMA provisions that govern fabrication.

The following includes some general design information that may be helpful to the designer.However, always refer to the above-listed guidelines and specifications for complete designand construction requirements of railroad underpasses.

♦ TxDOT will furnish no more than 16 ft.-6 in. of vertical clearance between the lowchord of a replacement or new construction railroad structure and the lower roadway.This is the usual maximum amount of vertical clearance allowed by the FederalHighway Administration (FHWA) for federally funded projects. Vertical clearance at asite where an existing railroad bridge is to be replaced will only be increased aboveexisting conditions as is economical and practicable for TxDOT. However, a limitedstudy by TxDOT has indicated that sufficient vertical clearance is the most importantfactor in guarding against over-height load hits. Therefore, special exceptions may beentertained on a case-by-case basis, provided they are negotiated before any designwork has begun and provisions for the railroad company to bear any additional costinvolved, such as additional railroad track approach work, have been agreed to in therailroad agreement and associated contract documents. Many times it is costly toprovide additional vertical clearance when replacing an existing railroad bridge becausethe line and grade of the existing track can be quite low and allowable grade on therailroad track can typically be less than 1 percent. Clearances for railroad underpassesare listed in Chapter 4 of this manual. Additional information can be found inChapter 4 of the Bridge Project Development Manual.

♦ TxDOT will not provide sacrificial beams ahead of the bridge that are not connected tothe bridge slab as a remedy for insufficient vertical clearance. This is based on theuncertainty of the applied load during an over-height load strike event and the ensuingpublic safety concerns. All beams in a railroad bridge will be placed at a similar low-chord elevation. Outside beams will not be placed at a lower low-chord elevation in aneffort to protect inside beams from an over-height load hit.



♦ All skewed railroad bridges must have an approach slab that squares up the edge of theconcrete deck with the railroad track to evenly support the first track tie on the bridge toprevent derailments caused by uneven track support. While a through girder structurehas the thinnest overall superstructure depth for a given span length it is also the mostcostly to build. A steel plate deck instead of a concrete deck may be used to decreaseoverall superstructure depth when vertical clearance is tight.

♦ Design is unusual for highway bridge designers because railroad live loads are so muchheavier. All underpasses are designed for Cooper E80 live load except for shoo-flybridges, which can be designed for Cooper E72 live load with railroad approval. Impactfactors for diesel locomotives are used exclusively, but they still vary with structureelement, type, and length of span.

♦ Because of a narrow composite flange and the magnitude of Cooper E80 loading, higherstrength concrete than TxDOT Class S must often be specified for slabs on prestressedconcrete or steel beams. Higher strength concrete than TxDOT Class C must be used ininterior bent and abutment caps because of the high beam bearing pressures underrailroad live load. In such cases, TxDOT Class F concrete with a minimum 28-daycompressive strength of 5,000 psi should be used.

♦ Beam spacing is always close and therefore deck forms often cannot be removed.

♦ A part of the Union Pacific Railroad Design Guide that deals with beam spacingsignificantly affects the choice of beam type for a railroad structure. Union PacificRailroad now requires a minimum of 18 in. clear distance between bottom beam flangesto facilitate beam and slab inspection. Spans up to about 95 ft. were achievable withAASHTO Type IV prestressed concrete beams spaced 1 in. clear under E80 loadingbefore Union Pacific Railroad instituted this requirement. The maximum span lengthnow attainable with AASHTO Type IV beams spaced 18 in. clear under E80 loading isabout 55 ft. At most grade separations, where TxDOT is replacing the railroadunderpass and widening the lower roadway, a span length greater than 55 ft. is requiredand, therefore, the superstructure must be carried on steel plate girders. However,Burlington Northern Railroad requires only 12 in. clear distance between bottom beamflanges in most cases and prestressed beams are still feasible for spans up to about 69 ft.under E80 loading for a Burlington Northern Railroad structure. Although BurlingtonNorthern Railroad does not have a published guide, it generally endorses the UnionPacific Railroad Design Guide; however, in many cases Burlington is not as stringent ifasked.


Salient features of the AREMA specifications affecting underpass design are:

Chapter 8 – Concrete Structures and Foundations

♦ Section 2 – Reinforced Concrete Design. Similar to AASHTO. Includes loadinggroups, and dead load factor of 1.4 is larger than AASHTO. Note the 25 percentlongitudinal force provision. Impact factors are usually larger than AASHTO and varyfrom member to member in the load path. Distribution of live load is very conservativeand can be open to interpretation by the various railroad companies. Lateral live load



distribution should be based on an 8 ft.-6 in. railroad tie. Minimum cover to bottomslab steel is larger. Minimum shear reinforcement is greater and proper detailing of laplengths for shear reinforcement in box beams requires careful attention. Yield strengthof reinforcement limited to 60 ksi.

Also includes Pier Protection (Railroad Crash Wall) Provisions. See text andcommentary for complete provisions.

♦ Section 5 – Retaining Walls and Abutments. The railroads can be very conservative intheir design approach for these elements, especially the distribution of live load throughearth fill. Methods vary for different railroad companies. Retaining wall abutmentsmust be designed for Cooper E80 surcharge with limited lateral distribution through thefill and without any support from the span. Mechanically stabilized earth retainingwalls are only acceptable for temporary installations.

♦ Section 16 – Reinforced Concrete Box Culverts. Contains design method for culvertsunder railroads. Live load is applied to the culvert regardless of depth of fill. Note the10 in. minimum slab and wall thickness requirement with 2 in. of cover over thereinforcement. Four percent minimum longitudinal reinforcement is required.

♦ Section 17 – Prestressed Concrete Design. Similar to AASHTO Section 9. Refers toSection 2 for loading and lateral live load distribution. No final tension in the concreteis allowed in prestressed concrete members. Calculation of impact factor forprestressed concrete is different than that for reinforced concrete members.

♦ Section 18 – Elastomeric Bridge Bearings. Very similar to AASHTO provisionsadopted in 1985. TxDOT’s experience and research tends toward a more conservativedesign, except allowable bearing pressure. If pressures greater than 1,000 psi areanticipated, approval should be obtained from the railroad company. Anchorageprovisions, one of which is the requirement for a 200 psi minimum dead load stress,requires that all bearing pads be doweled for a railroad bridge.

Chapter 15 – Steel Structures

♦ Section 1 – Design. Similar to AASHTO. Contains clearance, loading, and loaddistribution information. Load factor design is not allowed. Impact factor calculation isdifferent from reinforced or prestressed concrete design. Distribution of live load is veryconservative and can be open to interpretation by the various railroad companies.Lateral live load distribution should be based on an 8 ft.-6 in. railroad tie. Allowablefatigue stress ranges are the same as AASHTO for 2,000,000 and over 2,000,000 cyclesof loading, which are the only loading frequencies allowed. The mean impact fractionused for fatigue design is different than that used for strength design. AREMA“fracture critical” is the same as AASHTO “non-redundant.” Some railroad companiesuse a conservative interpretation of what constitutes a fracture critical member. UnionPacific requires an additional Cooper E65 loading check on the superstructure assuminga non-composite deck after designing the girders for Cooper E80 loading with acomposite deck. A useful table of simple span moments and shears due to Cooper E80loading is contained at the end of this section.



♦ Section 3 – Fabrication. The TxDOT Standard Specification refers to this section forthe control of railroad structural steel fabrication. Note the requirement for full shopassembly of a railroad bridge before shipment to the job site.

♦ Section 5 – Special Types of Construction. Section 5.1 covers composite design ofsimple structures. Section 5.2 contains a few instructions regarding continuousstructures; however, most railroads are reluctant to build them. There is no provisionfor composite continuous structures, but some have been used in the past. However,they are not recommended for use on TxDOT railroad projects due concerns mostrailroad companies have regarding their performance.

Chapter 19 – Bridge Bearings

♦ All bridge bearing provisions will be collected from the various chapters and includedin this section in the near future.

Chapter 28 – Clearances

♦ This chapter contains clearance diagrams for railroad bridge and track design.

Chapter 29 – Waterproofing

This chapter contains all specifications for waterproofing of railroad structures. Theseprovisions have been included in the Texas Standard Specifications (1993), Item 458“Waterproofing for Structures,” and the TxDOT plan details for Waterproofing and DeckDrainage. Use of the specification item and plan details will adequately cover theserequirements.

Crash Walls for Railroad Overpasses

Although the preceding section is about railroad underpasses, the need for crash walls is animportant aspect of railroad overpass design and deserves mention here.

Sufficient horizontal clearance between an interior support for a highway bridge and anunder-crossing railroad track must be provided or a crash wall is required to reduce thechance of the highway bridge collapsing after a hit from a shifted load on a train or a trainderailment. When the clear distance from the centerline of the track to the face of thecolumns is less than 25 ft. but greater than or equal to 12 ft., a crash wall extending 6 ft.above the top of the rail is required. When the clear distance from the centerline of the trackto the face of the columns is less than 12 ft., a crash wall extending 12 ft. above the top ofthe rail is required. Crash walls must extend 4 ft. below grade at the base of the wall and beat least 2 ft.-6 in. thick. Minimum reinforcing steel areas usually govern. See AREMAChapter 8, Section 2 for complete provisions and the Commentary for AREMA Chapter 8for a useful graphical representation of the provisions.

Refer to the Bridge Detailing Manual for detailing practice for crash walls. Horizontalclearance can not usually be less than 9 ft. unless specifically agreed upon by the railroad,usually in the case of an existing condition at a site.

Chapter 9 — Special Designs Section 6 — Pedestrian Underpasses


Section 6 Pedestrian Underpasses

Background

In the 1970s, approximately four pedestrian underpasses per year were constructed byTxDOT. Since 1980 there have been 19 on-system and 7 off-system pedestrian underpassesconstructed by TxDOT.

The following superstructure types have been used:

♦ Steel truss unit

♦ Steel arch unit

♦ Steel I-girder unit

♦ Steel I-beam unit

♦ Combination steel I-beam and plate girder unit

♦ Cast-in-place reinforced concrete tee girder unit

♦ Precast pretensioned concrete single tee beam span

♦ Precast pretensioned double tee beam span

♦ Cast-in-place reinforced concrete slab unit using both regular and lightweight concrete

♦ Precast pretensioned concrete beam spans – concrete deck on top

♦ Precast pretensioned concrete beam spans – concrete deck on the bottom flange of thebeam

♦ Precast pretensioned concrete beams – deck on top – continuous for live load

♦ Contractor-designed alternates (commercial designs)

More than half of these used precast pretensioned beams. Substructures have beenpredominantly single-column bents on drilled shaft foundations. Access to the underpasshas occasionally been by stairway, but mostly by tangent or circular ramps constructed withconcrete slab units.

Construction problems have been minor, except for one pedestrian underpass designed withcircular ramps composed of precast slabs resting on precast cantilever arms prestressed to acentral cast-in-place column. Twist and tolerance problems resulted in very difficultconstruction conditions.

One maintenance problem that should be remembered concerned a through prestressed beamunderpass with the deck slab resting on the lower flanges. The deck was cast to a levelgradeline and the beams sagged, with time, creating a pond of water on the walkway. It wasnecessary to drill holes in the deck to drain the water over traffic.



Current Status

Pedestrian underpasses are not popular because of long ramps required and a record ofinfrequent pedestrian traffic, but there is an increased interest due to TxDOT participatingmore heavily in city projects.


♦ Federal requirements, issued in 1979, require ramps to be designed to accommodatepeople who are disabled. Grades no steeper that 1:12 with level landings at 30 ft.spacing produce very long ramps.

♦ Vertical and horizontal clearances should be greater than for highway bridges inrecognition of the probability of severe damage from vehicle impact. Requiredminimum vertical clearance over traffic is 17 ft.-6 in.

♦ The usual width of walkway is 8 ft. Chain-link enclosure should be provided todiscourage foreign objects being dropped onto the highway. Enclosures should haverounded tops for aesthetics except that if any part is horizontally curved, the enclosureshould be flat-topped to avoid severe construction difficulties.

♦ Design live load is 85 pounds per square foot of deck, or one H5 truck.

♦ Reference the 1997 AASHTO Guide Specification for Design of Pedestrian Bridges foradditional considerations.

♦ Prestressed concrete beams, with cast-in-place concrete deck on top, are therecommended design. If vertical clearance is critical, the cast-in-place deck can beplaced on the lower flanges of the beams. This adds to the expense because of detailcomplication and the loss of composite action, but will still be more economical thanother minimum depth alternatives. Examples are shown in Figure 9-28 and Figure 9-29.

♦ Decks should not be constructed to a level gradeline. A minimum grade of 1 percent isrecommended as a straight grade or ending tangent to a vertical curve. Prestressedconcrete beam camber should not be considered to improve the drainage, but calculatedsag should be compensated by extra vertical curve or slope.

♦ A number of projects have been built recently using a one-time use special specificationfor a contractor-designed prefabricated pedestrian steel truss bridge. The contract plansprovide substructure details, which are paid for as traditional bid items, but thesuperstructure is bid as a lump sum item. The loading is the same and the minimumthickness of metal is 1/4 in.



Figure 9-28. : Typical Section of a Pedestrian Underpass – Example 1 (Online users canclick here to view this illustration in PDF.)



Figure 9-29. : Typical Section of a Pedestrian Underpass – Example 2 (Online users canclick here to view this illustration in PDF.)

Chapter 9 — Special Designs Section 7 — Historic Bridges


Section 7 Historic Bridges

Background

An increasing number of TxDOT projects involve “historic” bridges. Historic bridges aredefined as bridges older than 50 years that have been determined to be eligible for listing onthe National Register of Historic Places. The Environmental Affairs Division is responsiblefor making this determination of eligibility and for maintaining an inventory of historicallysignificant on-system and off-system bridges in Texas.

Current Status

The Historic Bridge Manual has been developed as a guide to planning projects involvinghistoric bridges. It provides a detailed explanation of the federal laws relevant to historicbridge preservation and sets forth procedures for project development that comply with thefederal preservation laws. These procedures are intended to minimize project delays thatcan occur as a result of the complex coordination that historic bridge projects require.


For design guidance refer to the Historic Bridge Manual.

Chapter 9 — Special Designs Section 8 — Long Span Bridges


Section 8 Long Span Bridges

Truss Spans

Background. Long span bridges of the past are simple trusses in Texas and the span lengthsare modest. Standard details can be found dating from 1918 for a variety of truss spans.

Pony trusses were used to approximately 80 ft. span. Most had riveted steel members, butone early standard had wooden chords and diagonals with steel verticals in the typical Howeconfiguration (see Figure 9-30). Most pony trusses were of the Warren configuration withverticals at even panel points. Wooden decks, used in the beginning, gave way to reinforcedconcrete in the middle 1920s.

Through trusses were used for 120 to 300 ft. spans. These were predominantly of the Parkerconfiguration with reinforced concrete decks (see Figure 9-31). Many had eyebars fordiagonals and lower chords. The last simple through truss span bridge was completed in1951. Through trusses are not compatible with high speed or high volume traffic, so manyof the old ones have been replaced. The "On-System Truss Bridges" table and the "Off-System Truss Bridges" table show the distribution of truss spans remaining on the statehighway system for on-system and off-system bridges, respectively. Many old trusses arelisted in the National Register of Historical Places, which makes replacement very difficult.

Design has been for service loads at working stresses according to classical textbookmethods. All steel trusses have been riveted, except for eyebars.

Fabrication was intensive because of the many relatively small and complicated members.Shop assembly before reaming field connection holes was required to assure the propercamber after erection.

Construction is complicated by the need for shoring during erection. In the latter days, rivetcrews were not easy to find. Proper closing of continuous trusses was sometimes tedious.

By far the most frequent maintenance required is the repair of portal bracing and trussmembers damaged by impact. They have been hit by overheight loads, overwidth loads,errant vehicles, and ships. In recent years, it has been necessary to replace some rusty rivetswith high-strength bolts.

Current Status. Trusses carrying highway loading are no longer constructed for the Texashighway system, but some of the old ones may be subject to Historical Bridge Restoration.

Design Recommendations. The AASHTO Guide Specification for Strength Design of TrussBridges2 was published in 1985. This would govern the design of trusses with 500 ft. orgreater spans. For shorter spans, service load design would be required in accordance withthe current AASHTO Specification.



Figure 9-30. : Configurations of Standard Half Through Trusses (Pony Trusses) (Onlineusers can click here to view this illustration in PDF.)



Figure 9-31. : Configurations of Standard Through Trusses (Online users can click here toview this illustration in PDF.)

On-System Truss BridgesConfiguration Combination

Deck/ThroughDeck Partial

ThroughThrough

Pratt Truss, Half Hip, Parallel Chord 2Warren Truss, Parallel Chord 3Continuous Truss 4Other Truss, Parallel Chord 1Pratt Truss, Parallel Chord 2Warren Truss, Parallel Chord 3Parker Truss, Polygonal Top Chord 21



Camelback Truss, Polygonal Top Chord 6Pennsylvania Truss, Polygonal Top Chord 1Other Truss, Parallel Chord 1Continuous Truss 4Other Truss, Parallel Chord 1Warren Truss, Parallel Chord 3Camelback Truss, Polygonal Top Chord 1Wichert Continuous Truss 1

Total 1 10 4 39

Off-System Truss BridgesConfiguration Combination

Deck/ThroughDeck Partial

ThroughThrough

Other than Metal Truss or Other Metal 1Pratt Truss, Parallel Chord 1Warren Truss, Parallel Chord 8Whipple Truss, Parallel Chord 1Camelback Truss, Polygonal Top Chord 2Warren Truss, Polygonal Top Chord 1Other Truss, Parallel Chord 3Other Truss, Polygonal Top Chord 2Pratt Truss, Parallel Chord 54Pratt Truss, Half Hip, Parallel Chord 2Warren Truss, Parallel Chord 9Warren Quadrangular Truss, Parallel Chord 1Whipple Truss, Parallel Chord 3Parker Truss, Polygonal Top Chord 6Camelback Truss, Polygonal Top Chord 4Pennsylvania Truss, Polygonal Top Chord 1Lenticular Truss, Polygonal Top Chord 1Pratt Truss, Parallel Chord 21Pratt Truss, Half Hip, Parallel Chord 10Warren Truss, Parallel Chord 380Bedstead Truss, Parallel Chord 4Parker Truss, Polygonal Top Chord 3Camelback Truss, Polygonal Top Chord 2Camelback Truss, Polygonal Top Chord 38Bowstring Truss, Polygonal Top Chord 2Lenticular Truss, Polygonal Top Chord 3Vierendell Truss 1Other Truss, Parallel Chord 2Pratt Truss, Parallel Chord 1Continuous Truss 2Other Truss, Parallel Chord 1



Total 1 22 466 81

Arch Spans

Background. The arch is one of the oldest means of framing an opening. Terms like“intrados,” “extrados,” and “springing” reflect the antiquity of this art form.

Early Concrete Arches. Early arch bridges in Texas were cast-in-place concrete. One of theearliest was completed in Austin with county funds around 1910, before the state highwaysystem was created. It was incorporated into the primary system in 1930, changed to a loopwhen the interstate highway was constructed, and taken off system again in 1986. Thebridge was widened in the 1950s without changing the basic structural elements. In the1970s distress was noted in the cantilevers from its spandrel walls. The city engaged aprivate engineering firm to design a wider superstructure while retaining the unloadedarches and spandrel walls for aesthetic and historical reasons. The new bridge wasconstructed with TxDOT supervision using federal and state funds. The result is consideredquite pleasing. It is now a roosting place for thousands of bats, which have become a vitalpart of the city’s image.

Other concrete arch bridges were constructed on the system until the late 1930s. Thoseconstructed under the National Recovery Act were especially ornate and of excellentworkmanship. An example of such workmanship can be seen in the Guadalupe RiverBridge in Comal County.

A similar bridge, completed in 1941 in Austin, marked the end of cast-in-place archconstruction in Texas. Rising labor costs had made the method uneconomical.

Steel Arches. Steel arches remained as aesthetic alternatives for longer spans. A steel I-girder, fixed end, deck arch span was constructed in the late 1950s as part of an interchangewith the Dallas-Ft. Worth Turnpike. Several tied arch bridges were constructed in the 1970sin other states. A steel box rib, part through, two hinge, 600 ft. arch span was designed bythe Bridge Design Section and completed across Lake Austin in 1982.

Design was interesting for the one steel arch span designed by the Bridge Design Section.The FHWA booklet on arch bridges3 was helpful in preliminary stages. The person whowrote this booklet was retained as an overview consultant, and was instrumental in changingthe original tied arch concept to the two-hinge design. A temporary hinge was provided atthe crown so forces would be determinate for the effects of rib weight. This eliminatederection stresses that would have been locked in due to placing the segments on falsework.

Analysis of the ribs was performed with the linear computer program FRAME 11 followedwith nonlinear FRAME 51.

Fabrication of the span was performed in South Korea with plant inspection by TxDOTpersonnel. There were many problems and much delay before the methods andworkmanship conformed to state specifications, but the finished product was of good



quality. Weathering steel conforming to A588 was used, and all shop connections werewelded.

Construction was accomplished without serious difficulty. Two false bents on steel pilingwere used for cantilever erection of each half span. Two barge-mounted cranes were used toerect most of the components. Field splices were bolted with A325, Type 3 bolts. Deck slabconcrete was pumped a considerable distance and deposited segmentally so as to load thearch ribs symmetrically.

No maintenance problems have been experienced on the Lake Austin bridge except for grassgrowing in the finger joint drainage troughs and graffiti on the arch rib. Some of the tiedarches in other states developed serious cracks in the tie. Wind-induced oscillation offloorbeam hangers was a problem in another state. Neoprene cable grippers were installedon the Austin bridge because of this, but no significant oscillation has been observed.

An innovative design was performed in the Houston District for four tied arch bridges tocarry city streets over U.S. 59, which was widened to add high occupancy vehicle lanes andshoulders. Single-span structures about 229 ft. long will have a depth of 1 ft. from roadwaysurface to the bottom of the tie, which allowed for additional vertical clearance under thestructures. The tie consists of a steel tube encased in concrete and post-tensioned to createno stress in the steel due to full dead load. This should eliminate cracking in the tie.

If this design fares well under traffic, it could find other applications, since space for bridgesis getting more critical.

Current Status. There are no arch bridges under consideration for Texas at the present time.

Design Recommendations. Avoid tied arches that require extra-high stresses in the tie.

A true arch, whether two-hinge, three-hinge, or fixed, delivers a large horizontal componentof reaction to the abutments. Anticipate expensive abutment design to withstand this thrust.

For the Austin bridge, load factor design methods were used for the arch ribs, with allowablestresses and plate thicknesses according to the AASHTO Specification. Analysis byFRAME 51 included secondary effects of deflection and rib shortening so that specifiedamplification factors were not required. Service load analysis, using FRAME 11, was usedfor deflections and stress ranges. Stress ranges due to live load can be severe, so carefulattention should be given to fatigue-resistant details and fracture-critical fabrication.

The Houston bridges were more complicated due to the construction sequence and post-tensioned tie. Computer programs STAAD 3 and StruCAD were used for analysis.

Cable-Stayed Spans

Background. Cable-stayed bridge concepts are said to have begun in the seventeenthcentury.4 Significant development for long span bridges did not occur until after World WarII in Europe. The first modern cable-stayed bridge was constructed in Sweden in 1955. Thefirst in the United States was completed in the early 1970s. Texas’ first was completed in1990 and its second in 1995.



In 1984 bids were taken for a bridge across the Neches River between Orange and PortArthur, Texas. Alternatives designed by the Bridge Design Section were provided for thechannel unit. One was a strutted steel I-girder unit and the other a balanced cantilever cast-in-place prestressed concrete box girder unit. Both had approximately 630 ft. effectivecenter span. According to the specifications, contractors could also submit an optionaldesign for any or all of the bridge. The lowest bidding contractor submitted a cable-stayedsegmental concrete box girder design for the channel crossing. The concept had beenapproved before bidding and the final design and details were approved later. Stays aregroups of prestressing strands in polyethylene sheaths anchored with strand chucks andgrout.

Texas’ second cable-stayed unit went to contract in 1986. It is located across the HoustonShip Channel near Baytown, Texas. Private engineering firms designed alternates, one withprestressed concrete box girder superstructure and the other with composite concretestructural steel I-girder superstructure. Center spans were 1,250 ft. The lowest biddingcontractor chose the structural steel superstructure. The contractor was allowed to changethe cast-in-place concrete deck slab to precast concrete. The stays are groups of prestressingstrands in polyethylene sheaths anchored with strand chucks and grout.

Design is very complicated. Negotiation of parameters was hampered by a lack ofauthoritative specifications. The PTI Recommendations for Stay Cable Design and Testing5

provided some relief. The American Segmental Bridge Institute (ASBI) has been formed toprovide a forum where owners, designers, constructors, and suppliers can meet to furtherrefine the procedure.

Both of these bridges have sustained damage in the anchorage zone of the stays. The secondbridge has severe oscillation of the stays in a mild wind. The first bridge has mildoscillation. A considerable amount of research by universities and forensic consultants hasbeen performed on the bridge stay systems since the bridges have been completed. Variousmethods such as cable dampening, cable restraining, and systems to change the aerodynamicresponse of the cables have been investigated, all with varying degrees of success.

Current Status. The Bridge Design Section does not recommend further use of cable-stayedbridges on the Texas highway system until proven systems are available to eliminate stayoscillations.

Design Recommendations. No design recommendations will be given.

Chapter 9 — Special Designs Section 9 — Bridge Railing


Section 9 Bridge Railing

Background

No component of modern bridges has undergone as many changes in design andconfiguration as bridge railing. Beginning with weak timber handrails or no railing at all andprogressing through massive concrete monstrosities, steel post railing of manyconfigurations, concrete posts with steel or concrete rails, concrete walls with aluminumrailing on top, concrete walls alone, aluminum post railing, and concrete safety shapes, theBridge Design Section has offered at least a hundred designs for use on Texas highwaybridges over the last 50 years. Individual districts added many other designs ormodifications to Bridge Design Section standards to satisfy their own conception ofappearance and functionality.

Current Status

For a complete discussion on the current status of bridge railing, refer to theBridge Railing Manual.

Retrofit Railing

Background. With the advent of the new railing designs, attention was directed to the manymiles of presumably unsafe railing already constructed on the Texas highway system. Aprogram was developed, and is still in progress, to make these old rails, of many and variedtypes, safer for the traveling public.

Current Status. For a complete discussion on the current status of retrofit railing, refer tothe Bridge Railing Manual.

Chapter 9 — Special Designs Section 10 — Expansion Joints


Section 10 Expansion Joints

Background

Bridge deck expansion joints have always been considered a necessary evil for bridges.They are problematic during design and construction, and require detailed attention whilebeing maintained throughout the life of the bridge. Due to these unattractive traits, attemptsare being made to minimize their usage. However, this too causes problems when improperprovisions are made to accommodate deck expansion.

Design of expansion joints has not been satisfactorily accomplished with analytical methods.Expansion joints are a product of ingenuity, which must be proved by tests and actualservice conditions.

During the past 20 years there have been many different expansion joint designs. Thefollowing are descriptions of some of the expansions joints that have been used:

♦ Early timber bridges could be designed jointless because of their resilience and loose fit.

♦ Early concrete slab spans, concrete girder spans, and some steel I-beam spans had pre-molded bituminous expansion joint material topped with poured asphalt in theexpansion joints. The poured asphalt was later replaced with a polysulfide compound.

♦ Truss spans with concrete deck usually had a plate attached to an embedded angle onone side of the joint and sliding over the top of an embedded angle on the other side.

♦ Steel I-beam details changed to open expansion joints, using embedded angles first,then vertical plates with anchors embedded in the deck.

♦ Continuous steel girder units, and later steel truss units, expanded through finger joints.Finger joints have been used for as much as 9 in. total movement. Numerousconfigurations of finger joints have been used, including those with drainage troughsmade with neoprene, stainless steel sheet metal, or heavy steel plate to direct the jointwater away from the bent. Refer to Figure 9-32 for an example of a finger joint.

♦ Prestressed concrete beams had small steel angles anchored with a single row of studswelded to the inside corner of the angles. Open steel plate armor joints with rows ofstuds soon replaced them.

♦ Neoprene compression seals known as preformed joint seals (PJS) were installed in thejoint opening in de-icing salt areas and in grade separation structures beginning in the1960s. Refer to Figure 9-33 for an example of an armor joint with PJS.

♦ A polyurethane compound joint seal (PUJS), which could be extruded or poured intothe armor joint opening, was used to a limited extent instead of neoprene compressionseals.

♦ A joint with a neoprene membrane clamped between plates and embedded angles wasdesigned in 1967 for use where leakage was particularly undesirable. This joint wasused extensively with various degrees of success. Its performance was highly sensitiveto the installation method. Details were changed several times to enhance performance.



However, the joint was removed as a standard in 1998. Refer to Figure 9-34 for anexample of this joint.

♦ Neoprene strip seals threaded or snapped into recesses in extruded or fabricated steelrails have been used for several years. The common name for this system is sealedexpansion joint (SEJ). The shape and anchorage of the rails has changed several timesover the years in an effort to enhance performance. Refer to Figure 9-35 for an exampleof a SEJ.

♦ Epoxy dams placed on top of concrete decks with asphaltic concrete overlay have beenused. The openings between dams were often filled with polyethylene rope and asilicone sealer to withstand small movements. This system usually failed miserably andshould not be used. The epoxy usually cracks and delaminates from the concrete due toa significant difference in coefficients of thermal expansion.

♦ Considerable research and testing has been done by TxDOT in developing a genericspecification for elastomeric concrete expansion dams. Several experimentalinstallations are being monitored. Strip seals are used in the joint openings.

♦ Small modular joints have been used, instead of finger joints, where required expansionexceeds the capacity of a 5 in. joint.

Construction Issues

Placement of expansion joints during construction is tedious. Some problems duringconstruction to be aware of include the following:

♦ The embedded parts of a joint must usually be supported securely while the concretedeck is placed and finished.

♦ Embedded anchor bolts for joints installed after the slab has been set may not fit.

♦ Anchors drilled into the slab are of questionable integrity.

♦ Large expansion joints may require anchor bolts and a slab block-out for properinstallation.

♦ Expansion joints must be carefully graded to avoid excessive roughness of the ridingsurface.

Maintenance Issues

Maintenance problems involving expansion joints have been numerous. Some examples ofthe maintenance problems encountered include the following:

♦ Unarmored concrete corners have spalled.

♦ Embedded anchors have broken.

♦ The open joint with small angles and one line of studs consistently malfunctioned.Rotation of the small angles under traffic was the probable cause.

♦ Sliding plate joints always slap under traffic and are difficult to repair.



♦ The top plates of membrane joints will become loose and dangerous if not carefullyinstalled.

♦ Aluminum joints with drilled anchors have consistently come loose under heavy traffic.

♦ Epoxy expansion dams often crack and separate from the deck.

♦ Neoprene expansion dams always lose their bolt hole caps and also tend to slap undertraffic. The latest problem involved the anchorage of strip seal rails. The rails had ahorizontal top flange with a line of 1/2 in. studs and another line of studs in the verticalleg. The studs broke under traffic. Postmortem examinations revealed voids betweenthe horizontal flange and the concrete due to poor concrete consolidation.

♦ Neoprene expansion dams have leaked between the elastomer and the top of theconcrete deck.

♦ Compression seals leak and often fall out completely because of irregularities in theopening, installation methods, or geometric problems. Polyurethane compounds provedunsatisfactory for continuous units because of cracking due to first night movements.

♦ Drainage troughs for finger joints become clogged with roadway debris.

♦ Membrane joints may leak at changes in geometry such as a skew break or turn up intoa concrete rail.

There is a serious effort underway, monitored by the TxDOT Bridge Division-BridgeConstruction and Maintenance Branch, to repair existing expansion joints that are leaking orotherwise malfunctioning. The most promising methods to date are the following:

♦ Asphaltic plug: A slab of rubberized asphaltic concrete placed over a 1/8 in. plateplaced above the open joint. Total movements of 1 1/2 in. have been accommodated.This method is not recommended for new construction.

♦ Elastomeric concrete or polymer nosing: A specially designed and constructed materialto rebuild spalled corners. Can be used with various sealing systems to waterproof thejoint.

♦ Class 7 silicone: A rapid curing sealing material that is placed on a foam filler insandblasted armor joints. Should be limited to 3 in. opening and a total movement of 50percent (+/-).



Figure 9-32. : Example of Finger Joint (Online users can click here to view this illustrationin PDF.)



Figure 9-33. : Example of Preformed Joint Sealer (Online users can click here to view thisillustration in PDF.)



Figure 9-34. : Example of Joint Type No Longer Used (Online users can click here to viewthis illustration in PDF.)



Figure 9-35. : Example of Sealed Expansion Joint (Online users can click here to view thisillustration in PDF.)



Current Status

Bridge deck continuity, which minimizes the number of expansion joints, is recommended.All expansion joints in the de-icing zones should be sealed or drained. Stream crossingstructures may have open armor joints in the salt-free zones. Joints for all grade separationstructures should be sealed.

The only membrane type joints approved for use are the following (see the TxDOT standarddetail sheets):

♦ SEJ-A, for 4 in. movement. The SEJ-A can be modified to allow a 5 in. seal. Refer tothe TxDOT standard detail sheet SEJ-A for the most current design details.

♦ SEJ-P, for up to 5 in. movement. The SEJ-P has a significantly heavier steel rail andshould only be used on bridges where heavy truck traffic is anticipated. Refer to theTxDOT standard detail sheet SEJ-P for the most current design details.

Finger joints should be used for joints requiring movement greater than 5 in.

Recommended joint types for all grade separations and for stream crossings in the de-icingzones are as follows:

♦ Pan form girder units:SEJ-A, armor joint sealed with silicone, or sealed elastomeric concrete

♦ Prestressed box beam units:SEJ-A, armor joint sealed with silicone, or sealed elastomeric concrete

♦ Prestressed concrete I-beam units:SEJ-A

♦ Steel girder units:SEJ-P for required movement capacity of 5.0 in. or lessInverted tee bent with two sealed expansion jointsFinger joints with drainage troughs for larger movements

♦ Concrete box girder units:SEJ-P for required movement of 5.0 in. or lessFinger joints with drainage troughs for larger movement

♦ Existing unsealed bridges:Bridges with ACP: Sealed elastomeric concrete or asphalt plug. Asphalt plugs haveservice limitations and should only be used after consulting with the BridgeConstruction and Maintenance Branch.Bridges without ACP: Blast clean and seal with silicone

Recommended joint types for stream crossings not in a de-icing zone are as follows:

Slab Spans and Units:Type A joint with preformed expansion joint material and silicone sealed top. One inchthickness is used for continuous units; 1/2 in. for simple spans.

♦ Pan form girder units: Open steel plate armor joints



♦ Prestressed box beam units: Open steel plate armor joints

♦ Prestressed concrete I-beam units: Open steel plate armor joints

♦ Steel girder units: Open steel plate armor joints or finger joints

The latest TxDOT standard detail sheets should be used for armor joints and sealedexpansion joints. Example details can be furnished for finger joints and troughs. Contactthe TxDOT Bridge Division-Bridge Construction and Maintenance Branch for details of theasphaltic plug.


♦ The total movement required through a bridge deck expansion joint may be based on120 degree temperature change (50 degree rise and 70 degree fall) for steel units and70 degree temperature change (30 degree rise and 40 degree fall) for concrete units.

♦ The joint opening at 70 degrees should be set near 0.42 times the total movementcapacity unless the recommended installation width of proprietary seals requiresotherwise.

♦ The maximum opening of the roadway surface, between steel armor plates, strip sealrails or membrane joint plates, should be 5 in. at the coldest design temperature.

♦ Anchors installed in drilled holes should not be used.

♦ Aluminum joint components are discouraged.

♦ Compression seals (PJS) and poured or extruded sealers (such as PUJS) are no longerrecommended for use. The TxDOT Bridge Division recommends a self-leveling lowmodulus silicone for joints with moderate anticipated movement, such as pan girders orslab spans. For joints with larger movements, a self-leveling, rapid curing, lowmodulus silicone is recommended. A new armor joint standard is being developed thatwill utilize the silicone instead of PJS.

♦ To reduce the need for large capacity joints such as finger or modular joints, an invertedtee bent cap with its stem extending through to finish grade has been used on a numberof projects; see Figure 9-36. The cap is constructed about 8 in. (10 in. if SEJ-P is used)below finish grade. A sealed expansion joint is placed at each face, and Class Sconcrete is placed with the last adjacent slab. Reinforcing steel strain in the upper pourdue to live load should be checked to control cracking. The corbel can usually bedetailed to make the bent aesthetically compatible with the other caps in the bridge.

♦ Finger joint design is subject to a variety of different support conditions. Teeth areusually sized for adequate flexural strength at working stress when loaded with theirproportional share of the load footprint. Tooth thickness to 2 in. is a nominal maximum.Stiffeners below the teeth are used when necessary to limit the thickness.

♦ Drainage troughs for finger joints are even more sensitive to configuration of structuralmembers. Nylon reinforced neoprene troughs accommodate joint movements well butare difficult to connect to structural members and downspouts. Metal scuppers attachedto one side of the joint with separate deflection baffles attached to the other side arepreferable, but often there is insufficient space to accommodate this. With either system



it is desirable to provide the maximum trough slope possible to minimize the probabilityof clogging by roadway debris.

Figure 9-36. : Example of Double Expansion Joints at Inverted Tee Bent Cap (Online userscan click here to view this illustration in PDF.)

Chapter 9 — Special Designs Section 11 — Bearings


Section 11 Bearings

Background

Bearings transfer loads from the superstructure to the substructure. There are two basictypes of bearings, fixed bearings and expansion bearings. Fixed bearings allow rotation ofthe superstructure but resist translation of the superstructure. Expansion bearings allow bothrotation and translation of the superstructure. A good state-of-the-art report on bridgebearings nationwide is given in Bridge Bearings, National Cooperative Highway ResearchProgram (NCHRP) Report 41 (1977). TxDOT has used many types of bearing mechanismsover the years. The following includes discussions on TxDOT’s experience with bridgebearings, and bridge bearing design, fabrication, construction, and maintenance history.

TxDOT’s experience with bridge bearings is as follows:

Timber spans began and ended with nothing between the stringer and cap but a vertical driftpin.

Concrete slab spans began with nothing at fixed bearings and tarpaper at expansionbearings. The tarpaper is now 30 pound roofing felt, with oil and powdered graphite to makeit slick. A strip of expansion joint material in the top corner of the cap allows a slightrotation of the spans without spalling the cap.

Concrete girder spans underwent the following evolution:

♦ Tarpaper

♦ Two layers of roofing felt separated by a copper sheet coated with graphite grease

♦ Cast steel curved plate sliding on a flat base plate

♦ Concave upper cast steel plate bearing on a convex lower plate with a 1/8 in. lead sheetbetween (for expansion, the bottom plate slid on a flat base plate)

♦ Welded steel plate bolster shoes with a pin connection for rotation and a rocker plate forexpansion

♦ Pan form girders started with roofing felt and still have it with oil, graphite, and capcorner protection

♦ The latest standard continuous concrete girders, fixed rigidly to the interior supports andslid across a convex steel plate at the ends

♦ A very few continuous concrete girders with lubricated bronze bearings

For steel I-beam spans the evolution was:

♦ Tar paper under a concrete end diaphragm

♦ Bolted to a flat base plate with slots in the beam for expansion

♦ Concave/convex cast steel plates with a lead sheet between



♦ Cast steel bolster shoes with continuous support on a pin and bolted to the beam

♦ Inverted convex fabricated steel plate with pintles to the flat sole plate

♦ Convex plate bearing on a base plate with anchor bolts through holes or expansion slotsin the beam flanges

♦ Bolster shoes fabricated by welding steel plates together (Rotation was allowed by a pintransferring the load through the vertical bolster plates. Expansion was accommodatedby a rocker.)

♦ A low profile end expansion rocker plate, pinned through the beam web with doublerplates

♦ Convex plate bearing on a base plate with anchor bolts clear of the beam flange

♦ Fabricated steel or cast ductile iron bolster shoes, welded to the beam

♦ A few of the later I-beam spans rest on plain or laminated elastomeric bearings

♦ End expansion bearings for a few continuous units used preformedfabric/Teflon/stainless steel sliding elastomeric bearings

Steel plate girders had mostly bolster shoes. Examples of bolster shoes are illustrated inFigure 9-37. Since the late 1960s many continuous units have been supported by laminatedelastomeric bearings at the interior bents and sliding elastomeric bearings at the ends. Referto the TxDOT standard detail sheet Elastomeric Bearing Details (For Steel Girders &Beams) for current standard design details. Lately, a few disc and pot bearings have beenused for steel plate girders as well.

The pin and hanger connection is a type of bearing that was used for a brief period in theearly 1940s. Most of them were on cantilever I-beam units. A few were used on non-redundant plate girder units.

Pony trusses tended to have flat plate bearings while deck and through trusses had mostlybolster bearings pinned to the lower chord connection. There is one roller nest installation ona long cantilever truss unit.



Figure 9-37. : Examples of Bolster Shoes (Online users can click here to view thisillustration in PDF.)

There have been about over 500,000 bearings for precast prestressed concrete beams used inTexas since the late 1950s. Except for a few steel bolster shoes, they have all been plain andlaminated elastomeric bearings.

Heavy concrete box girders, being constructed more frequently in the last few years, placedifficult demands on bearings. Usually, two bearings spaced close together on a capless



single-column bent carry the reaction of a wide roadway. Expansion bearings must allow forhorizontal movement of long continuous units and creep of the unit due to prestressing.Texas has used pot bearings, disc bearings, high load sliding elastomeric bearings, and largelaminated elastomeric bearings under these conditions.

Most of the steel trapezoidal box girders constructed to date have had pot bearings.

Bearings under steel box girder bent caps have been plain or laminated elastomeric andpreformed fabric.

Bearings under integral steel girder bent caps have been:

♦ Separated steel bolster shoes

♦ Single long steel bolster shoes

♦ Single long steel convex plate bearing on a base plate with prestressed anchor bolts

♦ Preformed fabric with high-strength anchor bolts

♦ Pot or disc bearings

♦ Separated shoes consisting of upright steel plates with convex top bearing in a concavedepression in a sole plate

This situation is usually severe because of the relatively close spacing of the bearings andlarge transverse overturning tendency due to off-center live load. One side will be heavilyloaded while the other side may try to uplift. Allowance for horizontal movement underthese conditions is impractical, but columns are usually sufficiently limber to accommodatethe deflection.

Anchor bolt strength is usually not critical for bridge bearings, but for one-column integralsteel bent caps careful consideration is necessary. Because threaded bolts are weak infatigue, the possibility of a significant tensile stress range has caused concern in the past andencouraged prestressing of the anchor bolts. Proper embedment may also severely limitanchor bolt strength.

Design Issues

The design of bearings involves prediction of the loads and movements to be sustained,followed by invention of a mechanism to accommodate them for a reasonable service of life.Bearings under deck joints are subject to severe conditions of dirt and moisture plus chlorideconcentration in the de-icing areas. A good bearing mechanism must survive thisenvironment. Some historical design experiences to be aware of are as follows:

♦ Design problems for steel shoes primarily have involved their ability to function underthe conditions experienced in the bridge. Analysis methods and allowable stresses areincidental. There was a long period of controversy over the allowable bearing stress forshoe pins. The 1973 AASHTO Specification settled it with a footnote stating that shoepins were “subject to rotation” and thus had a lower allowable bearing stress thanpinned truss members. Standard bolster shoe details issued in 1965 are slightly



overstressed in this respect, but no problems have been experienced. The method ofanalysis for base plates was usually debatable.

♦ Sliding elastomeric bearing design was justified by a few tests conducted by theMaterials and Test Division during a period when various manufacturers were trying todevelop Teflon/elastomeric combinations for bridge bearings. Under vertical load andhorizontal movement cycling, it was found that preformed fabric could restrain theflowing tendency of the Teflon, while other elastomers, without a metal interface, couldnot. Allowable design pressure was arbitrarily set at 1,000 psi, although the preformedfabric is capable of much higher loads. Performance of the Teflon interface would bequestionable under higher loads.

♦ Elastomeric bearings have undergone a controversial and largely empirical designdevelopment. In the late 1950s, neoprene bridge bearings had been used in Europe forseveral years and Dupont began to talk about this possibility in the United States.Dupont design data6 found its way into the AASHTO Specification in 1958, under thesection on expansion bearings for steel structures. By then, Texas was already usingneoprene bearings under prestressed concrete beam spans. Ten years later, elastomericbearings were given a section of their own in the specification, and natural rubber wasallowed as an alternative to neoprene.

♦ It was difficult getting neoprene accepted for bridge bearings initially. Then,controversy over the ability of natural rubber to perform acceptably and continuousarguments over cold temperature requirements delayed the 1968 specifications. Now,serious questions regarding the need for such a conservative design specificationcontinue to confuse the picture.

♦ NCHRP research7 in 1970 reminded bridge engineers of the variability of elastomericperformance, especially at cold temperatures. More extensive NCHRP research in 19828

led to a specification change in 1985 that the majority of Texas existing bearings couldnot meet because of shape factor limitations. Further NCHRP research9 proposed amore complicated empirical analysis method that allows higher bearing stresses. Roundrobin load tests performed in early 198010 revealed the ability of steel laminatedelastomeric bearings to withstand large compression loads. Further research has beenconducted by the Center for Transportation Research (CTR) with a focus on TxDOTpractice.

♦ Design of pot and disc bearings is largely based on manufacturers’ recommendationsfrom the results of their own research and testing.

Fabrication Issues

The quality of the manufactured bearing mechanism can affect a bearing’s performance asmuch as the design of the bearing. Some fabrication issues to be aware of include thefollowing:

♦ Fabrication problems with steel bearings concern tolerances and welding procedures forsteel plate bolster shoes.

♦ Most elastomer fabricators find it awkward to get another company to fabricate andbevel the steel sole plate for sliding elastomeric bearings.



♦ The formulation and fabrication of laminated bearings is very critical and complicated.The raw elastomer must be compounded with other ingredients to insure ozone andoxygen resistance and adhesion to laminated plates. All components must be kept cleanduring the manufacturing process and the mold must be designed correctly to preventseparation of the elastomer or misplacement of the laminate plates during vulcanization.

♦ For pot and disc bearings the condition of the steel backing behind the stainless steelsliding surface is critical because high loading can make the stainless steel assume theshape of the surface beneath.

Construction Issues

During bridge construction bearing mechanisms must be properly installed in order tofunction properly. Some construction issues to be aware of include the following:

♦ Construction problems with steel bearings concern getting an even bearing surfaceunder the base plate and welding the top bolster to the girder without distorting theflange.

♦ Sliding elastomeric bearings must be floated into wet grout to achieve lateral restraintfor the preformed fabric pad.

♦ Elastomeric bearings for prestressed beams are easy to construct due to a lack ofconnection between the cap and beam. This also makes them easy to mislocate, whichhas resulted in some bearings projecting outside the limits of the beam. This tends tocause splitting but not failure of the bearing.

Maintenance Issues

Properly maintaining bearing mechanisms also lends to their successful performance. Manytimes maintenance problems can be avoided if loading and environmental conditions can bepredicted and properly accounted for in the bridge design. Some maintenance issues to beaware of include the following:

♦ The older types of steel bearings have been the source of a few maintenance problems.Flat steel sliding plates freeze due to rust. Concave/convex surfaces work thelubricating lead sheet out by constant rotation and then the surfaces rust and freeze.Replacement of these shoes has been extensive. Bolster shoes have performed well,except for the short ones pinned through the webs of the I-beams. These are oftenunder leaky deck joints and are subject to rust build-up and freezing.

♦ Since the failure of the Mainus River Bridge in Connecticut, considerable attention hasbeen focused on pin and hanger connections. Texas had perhaps 20 or 30 bridges withpin and hanger connections of which only 3 or 4 could be considered non-redundant.Two of the redundant structures were found to have malfunctions of enough severity torepair. Pins were replaced on one of the non-redundant structures because of anultrasonic discontinuity in one of the pins.

♦ No problems have been reported with sliding elastomeric bearings.



♦ Problems with elastomeric bearings have often been caused by design deficiencies.Failure to anticipate the magnitude of horizontal movement has caused elastomericbearings to become sliding bearings. Slick bearings have contributed to bearingmigration. Plain bearings over one inch thick are subject to splitting along the bulgingedges, but no bearings have been replaced in the Texas highway system because ofsplitting.

In the middle 1980s, after almost 30 years use of elastomeric bearing pads and severalyears successful use of laminated bearings with their tops beveled to the slope of thebeam, TxDOT experienced a multiplicity of incidents of bearings migrating out fromunder prestressed concrete beams. This came at a time when the use of continuous unitsof as long as 400 ft. was prevalent, but the bearings were supposedly designed to notslip due to the extra movement. It was also a time when revised design specificationswere casting doubt on the wisdom of beveling the top of the bearing without providinga positive connection to the beam. After much study and testing and after reinstallingmany bearings, it was determined that the primary cause of the problem was a paraffinwax bloom on the surfaces of natural rubber bearings making them very slick. Thiswax is a necessary part of the formulation to satisfy current ozone resistance tests.Paraffin and microcrystalline waxes are mixed in varying proportions depending on thefabricator’s order, and as the amount of the less expensive paraffin componentincreases, so too does the amount of wax bloom. No one knew that bearings were beingmanufactured from natural rubber, nor had they attached any significance to the waxbloom until the migration problem was epidemic. This is the reason that natural rubberis now prohibited in TxDOT elastomeric bearings. As an added precaution, standardbearings were conservatively redesigned.

♦ All types of bearings have been punished by excessive substructure movement.Abutments regularly move toward the bridge due to fill settlement. End bearings can getcompletely extended and connections broken because of this. Some have been freed andstraightened, only to return to the same condition as the movement continues. Movingskewed abutments causes lateral distortion of bearings that cannot be accommodated bysteel bolster shoes. Interior bents have been displaced by movement of the riprap at thebridge end. Piers sometimes move along with the bank as it migrates toward the river.

This multiplicity of hardships makes it difficult for bridge bearings to survive.

Current Status

The following types of highway bridge bearings are recommended:Types A, B, and C prestressed concrete beams: Laminated elastomeric bearingsTypes 54, 72, IV, and VI(Mod) prestressed concretebeams:

Laminated elastomeric bearings

Prestressed concrete U-beams: Laminated elastomeric bearingsSteel I-beam spans: Laminated elastomeric bearingsCurved and tangent closely spaced steel girder units: Laminated elastomeric bearings for interior reactions

Sliding elastomeric bearings for end reactionsWidely spaced steel girder units and steel trapezoidalbox girder units:

Pot or disc bearings, laminated elastomeric bearings

Simple and continuous reinforced concrete slab spans 1/8 in. asphalt board with oil and graphite at expansion



and pan form girder spans: ends, with cap corner protectionDoweled at fixed ends, with cap corner protection (Referto Figure 7-3 and Figure 7-4 for examples.)

Prestressed concrete slab units: Asphalt board, oil, and graphite at expansion end, withcap corner protectionDoweled at fixed joints with neoprene ring columncorner protection (Refer to Figure 7-12 for an example.)

TxDOT box beam spans: Laminated elastomeric bearings, two at one end and oneat the other

Concrete double tee spans: Laminated elastomeric bearingCast-in-place concrete box girder: Rigidly fixed to interior supports

Sliding elastomeric bearings at expansion endsPrecast segmental concrete box girders: Preformed fabric bearings at fixed supports

High load sliding elastomeric bearings at expansionsupports

Steel box girder bent caps: Preformed fabric bearingsIntegral steel girder bent caps: Separate shoes with upright convex plate bearing in

concave depression in a sole plate (Refer to Figure 9-38and Figure 9-39.)

Pin and hanger connections should not be used in non-redundant structures.



Figure 9-38. : Example of Bearings for Integral Steel Girder Bent Caps – View 1 (Onlineusers can click here to view this illustration in PDF.)



Figure 9-39. : Example of Bearings for Integral Steel Girder Bent Caps – View 2 (Onlineusers can click here to view this illustration in PDF.)


Suggestions will be given for steel components, elastomeric bearings,polytetrafluoroethylene surfaces, preformed fabric pads, and pot and disc bearings.

Steel Components



Steel components occur in most types of bearings. Steel plate should usually conform toAmerican Society of Testing and Materials (ASTM) A709, Grade 36, or Grade 50, and itshould be painted if exposed. If weathering steel is used, details that promote a continuousmoist condition must be avoided. Pins may conform to ASTM A108 (4 in. or less indiameter) or ASTM A668 Class C, D, F, or G, as required by AASHTO and the TxDOTconstruction specifications. Pins get a special coating of zinc oxide, tallow, and linseed oil.

♦ Allowable bearing stress on pins is 0.4 Fy (yield stress of pin or bearing plate).

♦ Allowable bending stress in pins is 0.8 Fy.

♦ Pins and pin holes are finished to ANSI 125 roughness or better.

♦ Allowable bearing stress on rockers, in pounds per linear inch, is:Fy 13000

20000 600d−� �

OR

Fy 1300020000 3000 d−�

��

��

, for 25 in. ≤ d ≤ 125 in.

Where:d = diameter of rocker or roller in inchesFy = minimum yield point in tension of steel in the roller or bearing plate,Whichever is the smaller.

♦ Rockers are finished to ANSI 250 roughness.

♦ Base plates and sole plates bearing on concrete should be sized for a maximum pressureof 0.3 f 'c, increased by the square root of the plate area divided into the area ofconcentric concrete, not to exceed 2. Plate thickness should be based on service loaddesign, using the bearing pressure as load with supports as defined by the connectingcomponents. Concrete bearing areas should be reinforced to resist bursting stresses.

♦ Unless otherwise required, the components should be capable of transmitting ahorizontal force of one-tenth of the vertical capacity from superstructure to substructureat working stresses.

Rotation of 0.02 radians (0.02 ft. per ft.) may be assumed for dead load plus live loaddeflections.

Pin and hanger connections should conform to the AASHTO Specification for “Links andHangers” and the following:

♦ Hangers should be constant width plates of A709, Grade 50 steel with maximum widthof eight times their thickness.

♦ Hanger plate area should have a tensile working strength, on the net section through thepin, of at least 1.4 times the calculated reaction due to all loads.

♦ Pins should be ASTM A668, Class F or G with recessed pin nuts.



♦ The allowable shearing stress in pins and the allowable bearing stress of pins on hangersand reinforced girder webs should be 20 ksi.

♦ Plates reinforcing the girder web for bearing stress should be ASTM A709, Grade 50,symmetrical either side of the web and fully developed by fillet welds and possibly plugwelds in shear. Plug welds should be used when the distance between attachment wouldotherwise exceed 12 times the thickness of reinforcing plate.

♦ Buckling of the web plates should be investigated.

Elastomeric Bearings

Elastomeric bearings shall be laminated neoprene (polychloroprene). Solid steel sheet orplate is used for laminates, usually 12 gauge 0.105 in. thick. Steel “sole” plates may bevulcanized to the bearing for attachment. Each laminated bearing must be molded andvulcanized separately. Mold cost is significant if only a few bearings of that size will berequired. Standardization of sizes is preferred, but mold costs should not preclude designingnew shapes if the situation demands. There is a limit to the size of bearings that can bevulcanized, depending on the size of press available for containing the mold and elastomerduring vulcanization. Sizes greater than 42 in. square should not be specified without priorconsultation with a reputable bearing manufacturer.

Accurate design of elastomeric bearings is virtually impossible. The elastomer is aproprietary mixture of ingredients that must be loaded into the mold with extreme care andcured under high temperature and pressure to attain a measure of integrity. A tolerance of0.005 radians is allowed for deviation of the bearing surface from the theoretical plane. Thebearing area on the bent has a tolerance for flatness. The beam surface that rests on thebearing has another tolerance in addition to its variation due to camber and deflection. If allof these tolerances were to accumulate on one corner, a sizable compressive strain would beinduced in an ordinary bearing thickness. There is an amount of creep equal to 25 percent+/− of initial dead load deflection, under sustained load. Initial deflection under dead load isusually around 4 percent for 50 durometer constant thickness laminated pads with shapefactors of approximately 10.0. Tapered bearings deflect up to 60 percent more than constantthickness pads, for tapers approaching 0.06 radians. In the structure, bearings are subjectedto fluctuation in vertical and horizontal movement and rotation at extremes of ambienttemperature, which significantly influence elastomer performance. There is evidence thatfluctuation of vertical load and horizontal movement can cause fatigue damage to thebearing. However, this fatigue damage is a rare occurrence. Controversy over the latestdesign specification is understandable. TxDOT elastomeric bearing design practice departsfrom the AASHTO design specification in several key areas. Taper, allowable compressivestress, cold temperature shear modulus, rotation, stability, and testing procedures formaterial properties are all areas that TxDOT has chosen to base its design philosophy onresearch results as well as on extensive field experience. Use of AASHTO methodologyalone will result in usage of bearings from TxDOT standard sheets that is in conflict withintended applicability.

Elastomeric bearings for Type A, B, C, IV, 54, and 72 beams and utilizing various beamangles have been standardized. Details can be found in the TxDOT standard detail sheetElastomeric Bearing and Beam End Details.



Type VI(Mod) beams require laminated 50 durometer bearings of special design.

These bearings are considered adequate for use with three-span units made up of equallength spans of a given beam type. Units made up of Type A or Type B beams may use thestandard pad on up to four equal length spans. Standard bearing pads to be used as endbearings on any unit with three or more spans should be checked for compliance to slipprevention equations. Short end spans, narrow beam spacings, and severe grades can makethe standard bearing pad unacceptable for use on the ends of multiple span units. Strainrequirements will be satisfied by virtue of restricting the unit length to less than 400 ft. Atthe ends of longer units with continuous prestressed concrete beams, special bearings shouldbe designed according to the following provisions for prestressed concrete beams.

♦ For roadway grades equal to 5 1/2 percent or less, these bearings should have the topsurface beveled to match the slope of a line between the bottom of the beam elevationsat each end of the beam. For roadway grades greater than 5 1/2 percent, specialanchorage details may be required. Examples of special anchorage include fixed beamends at low end, level beam bearing surfaces, sole plates, etc.

♦ Various modifications to the plan configurations, including round bearings, to fitstandard bents at different skew angles are shown on TxDOT standard drawing IBB.Note that Types IV and 72 require special consideration of the bent cap width.

♦ For other elastomeric bearings, the provisions of Method B of NCHRP Report 325 andnow the AASHTO Specifications are recommended, with clarifications andmodifications given below:

Note: The shearing modulus of elasticity (“G”) may be taken as follows:

Shearing Modulus of ElasticityHardness At 70°°°° F At 0°°°° F

50 Durometer 110 psi 175 psi

Note: Test results are inconsistent. These values are considered sufficiently adequate forTxDOT bearing design. The 70° F value is used for allowable compressive stress. The0° F value is used for slip calculations.

♦ Compressive stress/strain performance of elastomeric bearings may be taken fromFigure 9-40.

Note: There is a wide variation in these properties among available sources ofinformation. These graphs were taken from two sources.11 Also, as previously noted,tapered pads will deflect more than constant thickness pads. Figure 9-40 givesdeflection characteristics for constant thickness pads.

♦ For elastomeric layers thicker than 3/8 in., the thickness of the outer layers of elastomershall be no more than three-fourths of the thickness of the inner layers.

♦ The total elastomer thickness (all layers) shall be no less than twice the calculatedtranslation in one direction based on an assumed temperature fluctuation of 70° F fromthe installation temperature.

Note: This allows a shear strain of 50 percent.



♦ Translation of bearings (from installation position) may be calculated from thefollowing strains:Temperature, Concrete 6.0E-06 x 70 ° F x 12 = 0.0050 in./ft.Temperature, Steel 6.5E-06 x 70 ° F x 12 = 0.0055 in./ft.Shrinkage, Concrete 0.0002 x 12 = 0.0024 in./ft.Elastic Shortening, Prestressed Concrete = 0.0024 in./ft.Creep, Prestressed Concrete 0.0005 x 12 = 0.0060 in./ft.

Note: For the typical prestressed concrete beam pad design, shrinkage, elasticshortening, and creep need not be taken into account. A “one slip” philosophy, wherebyall movement that occurs after beam erection, other than that caused by thermalcharacteristics, may be disregarded due to the beam slipping on the pad one time andthereby restoring the pad to an equilibrium position. From this equilibrium position, thepad will accommodate thermally induced shear translation in either direction for theduration of the structure’s life.

♦ The use of adhesives between the bearing and beams or bearing seats is notrecommended.

Note: Bearings have been observed to migrate because adhesives are broken by flexingof the elastomer, leaving a very slick interface.

♦ For precast prestressed concrete beams and steel girders these requirements may besummarized as follows:

T = Total elastomer thickness (in.)t = Interior layer thickness (in.)L = Bearing length across beam (in.)W = Bearing length along beam (in.)D = Diameter of round bearings (in.)SF = Shape factor:Rectangular — L x W/(2(L+W)t)Round — D/4A = Area of bearing (in.2)Rd = Dead load reaction (lbs.)Rt = Total reaction (lbs.)

L' = Expanding length (ft.)F = Dead load reduction factor due to beamslope or grade = (0.2-Gr)/0.2Gr = Beam slope (ft./ft.)G = Shear modulusHardness = 50 DurometerMinimum shape factor = 6Target shape factor = 10 to 12Maximum Rd/A = 1,200 psi, or 1.2 x G x SF,whichever is lessMaximum Rt/A = 1,500 psi

Note: This is a simplification of portions of Method B of the AASHTO Specificationsin view of the demonstrated ability of laminated elastomeric bearings to sustain muchhigher loads. The provisions for simultaneous compression and rotation have beenrejected because of undue complication.

♦ For Translation (maximum shear strain):• Minimum T = 0.0055(2)(L') = 0.011 L' (Steel)• Minimum T = 0.0050(2)(L') = 0.010 L' (Concrete)

♦ For No Slippage (without anchorage):• (Steel) Minimum T = 0.005 (L')(175)(A)/[(0.2)(Rd)(F)]• (Steel) Minimum T = 4.8 L' A / [(Rd)(F)]• (Concrete) Minimum T = 0.0050(L')(175)(A)/[(0.2)(Rd)(F)]



• (Concrete) Minimum T = 4.4 L' A / [(Rd)(F)]• Maximum T = the lesser of L/3 or W/3; for rectangular bearings• Maximum T = D/4; for round bearings

Note: For stability, the provisions of Method B have been rejected because of theirunnecessary complication.

♦ If upper and lower bearing surfaces are cast against the in-place bearing, the only non-parallelism that needs to be considered is longitudinal rotation due to the dead load andlive load deflection. This may be taken as 0.01 radians.

Caution: There are no construction tolerances for this method.

♦ If upper and lower bearing surfaces are precast, non-parallelism may be assumed to be0.02 radians longitudinally.

Note: This includes a nominal amount for construction tolerances and live load rotation.If all of the allowable construction tolerances were to occur in the right direction toeffect one corner of the bearing, the compressive strain on that corner would be severeand there might be daylight between beam and bearing on the other side. Thesenumbers are a judgmental recognition of the problem. Transverse non-parallelism waseliminated from consideration in the Method B Specification.

Note: To keep the beam from lifting off the pad by more than 20 percent, a deflectiondue to rotation equal to 0.02 x (W x 0.8) / 2 should not exceed the compressivedeflection due to total load. The compressive deflection can be estimated by using theperformance curves in Figure 9-40 with Rt /A and SF to read compressive strain. Thisstrain is then multiplied by T to obtain the estimated deflection.

Note: Research has indicated that the pad will continue to function properly with up to20 percent of lift off (a departure from AASHTO design philosophy for rotation), ergothe “0.8” factor applied to the pad “W” value.

Note: Higher shape factors call for increased thickness if this control is to be observed.

♦ For tapered pads, the estimated compressive deflection (as determined from theperformance curves) should be increased by 10 percent for every 0.01 radians of beamslope. This will increase the allowable rotation in certain cases. For typical prestressedconcrete beam spans, the total deflection is usually less than 3/16 in., and thereforerarely a consideration where differential deck surface elevations are concerned. Thickerthan standard pads with smaller shape factors and moderate to extreme tapers, maymerit checking for maximum compressive deflection by the designer.

♦ If total deflections are needed, multiply the above by 1.25 for creep.

Sliding Elastomeric Bearings

Sliding elastomeric bearings contain preformed fabric, as defined in the constructionspecification, for the element that allows rotation. The specification should also referencemilitary specification MIL-C-882D. Typical bearings are shown in the TxDOT standard



detail sheet Elastomeric Bearing Details (For Steel Girders & Beams). Salient features ofthe design are:

♦ The preformed fabric should be 2 in. thick.

♦ The average bearing pressure due to dead load and live load without impact should notexceed 1,100 psi.

♦ Preformed fabric pads should be laterally restrained on the bearing seat as shown in theTxDOT standard detail sheet Elastomeric Bearing Details (For Steel Girders & Beams)or by other mechanical means. Adhesive should not be used.

♦ Bearing seats should be level; sole plates should be beveled to the slope of the beamseat adjusted for dead load rotation.

High load sliding elastomeric bearings have been specified for concrete box girders.Preformed fabric is bonded to a steel plate that is recessed to receive a PTFE sliding surface.The top is stainless steel on a steel backing plate. Anchorage to the upper and lower concretesurfaces is with end welded studs. Under these conditions, the following controls arerecommended:

♦ The preformed fabric should be 2.0 in. thick.

♦ The average bearing pressure on the preformed fabric due to dead load and live loadwithout impact should not exceed 2,000 psi.

♦ The average bearing pressure on the PTFE surface should not exceed 3,000 psi.

♦ Top and bottom surfaces should be level. Concrete or epoxy grout should be castagainst the in-place bearing.

Pot and Disc Bearings

Pot and disc bearings will usually have design parameters defined in the constructionspecification. As currently written these parameters are as follows:

♦ Minimum horizontal capacity is 10 percent of vertical capacity.

♦ Minimum rotation capacity is 0.02 radians.

♦ Allowable stress for steel and concrete is as prescribed by AASHTO.

♦ Allowable average compressive stress on elastomeric discs and PTFE surfaces is 3,500psi.

♦ Minimum thickness of plate beneath pot or disc is 0.06 D.

♦ Minimum thickness of elastomeric disc is 0.07 D.

♦ Top of pot disc to top of pot walls is 0.015 D + 0.12 in.

♦ Minimum pot wall thickness is 0.75 in.

♦ Minimum depth of limiting ring for disc bearing is 0.014 D.

♦ Bearings are to be designed so that all rotational and sliding elements can be replacedwith a minimum of jacking.



Figure 9-40. : Elastomeric Bearing Performance (Online users can click here to view thisillustration in PDF.)

Chapter 9 — Special Designs Section 12 — Anchor Bolts


Section 12 Anchor Bolts

Background

Many anchor bolts have been used in Texas for various purposes, amid considerablecontroversy regarding materials and embedment addressed by numerous research projectsover a 20 year period.12

The AASHTO Specifications offer little guidance, other than a few antique rules forminimum size and embedment for bearings under steel beams, girders, and trusses.

Anchor bolts for steel bearings are nominally designed. Anchorage has been obtained byusing headed bolts, swedged bolts, or threaded rods with bent ends. For the usual beambearing, anchor bolt tension is impossible, short of a catastrophe, and shear requirements aresmall. Shoes are usually painted but anchor bolts have vacillated between painting andgalvanizing depending on the construction specifications of the period.

The advent of large roadside signs and overhead sign bridges created a need for seriousconsideration of tension in anchor bolts. Texas’ earliest research13 was directed to thisproblem.

The new specification for bridge railing called for smaller high-strength bolts with differentanchorage conditions. Galvanized A321 threaded rods with nuts or A325 headed bolts wereused almost exclusively for this. Anchorage was usually sufficiently strong to break theconcrete slab or parapet when severely impacted. Some of these railing systems and manyolder railings are now being replaced, often requiring anchor bolts to be installed in existingconcrete. Design Guide for Steel to Concrete Connections, by Cook, R.A. and others (CTR,Report 1126-4F, 1990) provides means of determining strength of several types of retrofitanchors.

Complicated framing of steel cap beams required for geometrically restricted urbanfreeways often placed severe theoretical stresses in anchor bolts embedded in concretecolumns. This instigated the research reported in Strength and Behavior of Anchor BoltsEmbedded Near Edges of Concrete Piers by Hasselwander, G.B and others (CFHR, FinalReport 29-2F, 1977). High theoretical tensile stress ranges discovered in one interchangeafter construction prompted some undocumented load testing and stress measurement.Although the theoretical stress ranges were not apparent during the testing, preformed fabricwashers were installed to reduce the stress ranges to insignificant values. Laboratory tests, inconjunction with this problem, confirmed the findings of Axial Tension Fatigue Strength ofAnchor Bolts by Fischer, F.L. and K.H. Frank (CFHR, Report 172-1, 1977) that even high-strength threaded anchor bolts have low fatigue strengths.

Stress range in anchor bolts for cantilever sign bridges became a concern because of thenoticeable traffic-induced deflections. Fatigue Loading in Sign Structures by Creamer, B.H.and others (CFHR, Report 209-1F, 1979) reported that the frequency of deflection was veryhigh but the stress range induced was well below the threshold level for threaded bolts.



Early anchor bolts were A307 or A36 threaded rods. High-strength bolts required for signsupports were A325 or A321 threaded rods, for which availability was questionable for thelarger sizes. A search for the best available high-strength bolt material identified A193,Grade B7, a chrome-molybdenum steel developed for pressure vessels, as being readilyavailable at reasonable cost. Tall illumination poles introduced the Grade 75 deformedreinforcing bar, size 18S, with cut threads for anchor bolts. Threading did not remove all ofthe deformations, which created concern about strength of the threads. Rolled threads wererequired for awhile, but debate ensued over the relative fatigue strength of rolled threads andcut threads. Axial Tension Fatigue Strength of Anchor Bolts by Fischer, F.L. and K.H. Frank(CFHR, Report 172-1, 1977) reported little difference.

Concern over fatigue occasionally precipitates a prestressed anchor bolt design. This workswell when regular prestressing materials are used. Attempts to prestress A193 anchor boltsfor tall light poles were complicated, had questionable results, and have been abandoned.Fatigue of Anchor Bolts by Frank, K.H. (CFHR, Final Report 172-2F, 1978) reported somefatigue benefit from local bolt prestressing within the thickness of the base plate provided bydouble nuts top and bottom of the plate.

Because of fatigue concerns, the allowable stress in anchor bolts for sign support and talllight poles was limited to the specification factor times 55 ksi yield stress. This limited theeffectiveness of Grade 75 reinforcing bars, A321 threaded rods, and A193 material.Manufacturers have developed a 55 ksi bolt material, called A36M55, in response to thisrequirement. Scratch gauge strain measurements made by the Bridge Design Section severalyears ago revealed no significant stress ranges in tall light poles, but anchorage strength nowappears to limit the usefulness of higher strength steels.

Construction Issues

Construction problems are caused primarily by mislocated anchor bolts. For bridgesuperstructures, anchor bolts are often set in preformed holes in the concrete to allow forvariation in bent locations or beam length. Templates are a virtual necessity for anchor boltgroups, as for sign supports and tall poles. Prestressed anchor bolts present an addedconstruction difficulty. Coordination between traffic signal pole design and anchor bolt sizeswas a big problem before standards for these poles were completed.

Maintenance Issues

Maintenance has been largely confined to rusty bolts in shoes for steel beam units. In spiteof all the concern over embedment and fatigue, no problems have been manifest, except forone cantilever sign bridge that fell onto an interstate freeway in Houston. The anchor boltshad fractures that indicated fatigue, but the structure had been modified in a manner thatincreased the theoretical bolt stresses considerably beyond specification allowables. Thus,the fatigue was probably caused by a small stress range from a resident tension approachingyield stress.



Current Status

Current usage of anchor bolt materials is as follows:

♦ ASTM A307, A36:• Bearings for redundant superstructure beams• Pedestrian and bicycle railing and T6 traffic railing

♦ A36M55, F1554 Grade 55:• Traffic signal poles• Overhead sign bridges• High mast illumination poles• Luminaire poles

♦ ASTM A321, A325:• Posts for traffic and combination railing

♦ ASTM A193-B7, A687:• Overhead sign bridges• High mast illumination poles• Luminaire poles• Steel girder bent caps

Anchor bolts are covered under a separate item in the current TxDOT constructionspecifications. Four types are identified, as follows:

♦ Mild steel anchor bolts:• ASTM A307 Grade A or ASTM A36

♦ Medium strength, mild steel anchor bolts:• ASTM A36 or ASTM A572 with 55 ksi minimum yield stress, called A36M55

♦ High strength anchor bolts:• ASTM A325 or A321

♦ Alloy steel anchor bolts:• ASTM A193-B7 or ASTM A687

Suitable nuts and threads are specified for each type.


Recommended service load stress allowables for the four types of anchor bolts are asfollows:

♦ Mild steel (Minimum Fy = 36 ksi)Tension 18 ksi



Shear 11 ksiFatigue 8 ksi

♦ Medium strength mild steel (Minimum Fy = 55 ksi)Tension 27 ksiShear 16 ksiFatigue 8 ksi

♦ High strength (Minimum Fy = 70 ksi to 2-1/2 in. Dia.)Tension 36 ksiShear 21 ksiFatigue 8 ksi

♦ Alloy steel (Minimum Fy = 105 ksi to 2-1/2 in. Dia.)Tension 50 ksiShear 30 ksiFatigue 8 ksi

The allowable tension for high-strength and alloy steel bolts will probably be unachievabledue to fatigue or anchorage limitations. Sign bridge and light pole anchor bolts have beenlimited to a design tension of 27.5 ksi and shear of 16.5 ksi, for these reasons, althoughA36M55 or ASTM A193 bolts are required. Fatigue allowables are based on Category E,redundant for 2,000,000 cycles. The stress area, through the threads, should be used in allcases. Allowables may be increased for certain load groups as allowed by AASHTO bridgeor sign bridge specifications.

Shear due to torsion in the anchored member should be added to the shear caused bytransverse forces. Bending stresses due to unsupported anchor bolt length may be ignored ifprojection is no more than five times the bolt diameter and double nuts are securelytightened.

Specifications are unclear regarding the combination of tension and shear in anchor bolts. Itappears prudent to utilize a method similar to the requirements for connection bolts so that:

( ) 22tension0.6

shear +�

��

�

�

≤ Allowable Tension

Allowable tension should be commensurate with anchorage strength. Most authoritiesemphasize the desirability of allowing “ductile” failure by sufficient anchorage strength todevelop the yield or even the ultimate strength of the bolts. This is a virtual impossibility foranchor bolts in columns or parapet walls such as commonly required for sign bridges, lightpoles, and bridge railing. Certainly, the anchorage should develop the service load tension inthe bolts. Preferably, a reasonable factor of safety should be provided.

Several reports14 explored several factors that affect anchorage strength. Nuts and washers(or plates) were found to be the most effective means of anchorage. Clear concrete coverand anchor bolt spacing are limiting factors. Research recommendations were as follows:



Tn = 140 Ab f 'c 0.7 ln 2CDw D Ks+ −� �

�

��

��

where:Tn = nominal tensile capacity of a boltAb = net bearing area, in.2, calculates as �

��

� − 22 DDw4π ,

and not greater than 4D2

D = bolt diameter, in.Dw = the diameter, in., of the washer or anchor plate; where a continuous template or anchor

plate is used for a group of anchor bolts, the washer diameter may be taken as the diameterof a circle concentric with the bolt and inscribed within the template or anchor plate. Dwshall not be taken grater than eight times the thickness of the washer, plate, or template.

C = Clear cover to bolt, in.Ks = Spacing reduction factor = (0.02S + 0.40) < 1.0S = Center-to-center bolt spacing, in.f 'c = Concrete compressive strength, psi

Embedment must be at least 12(Dw − D) to ensure wedge splitting failure. Additionally,local practice requires a minimum embedment of 20 bolt diameters.

These will be recognized as stringent limits on anchor bolt effectiveness. The designer is leftto individual judgment in the application of load and resistance factors. The researchersrecommended a resistance factor of 0.75 times Tn as defined above. It is comfortable to notethat there have been no anchor bolt failures in Texas except for rust and vehicle impact.

Strength and Behavior of Bolt Installations Anchored in Concrete Piers by Jirsa, J.O. andothers (CTR, Final Report 305-1F, 1984) appears to offer a more comprehensive method todesign the anchorage to develop ultimate strength of the bolts.

It is recommended that all anchor bolts with calculated tension, except for railing anchorage,be anchored with embedded nuts and plates that can serve as a template as well as increasethe pullout resistance. Nuts should be tack welded to the template to prevent floating duringconcrete placement. Anchor bolts for bearings that resist no uplift may be anchored by 12 in.embedment and a 2 in. right angle bend.

Anchorage zone reinforcing should be given careful attention. Supporting members must beable to resist the forces delivered by the anchor bolts. Vertical reinforcing must havesufficient development length to resist stresses delivered by the anchorage plate.Confinement reinforcing is beneficial, and transverse reinforcing can improve the shearstrength of the anchorage.

For severe tension or fatigue conditions, consideration should be given to prestressedanchorage. Regular post-tensioned systems should be used. Prestressing anchor bolts byturning the nut is of doubtful value.

The strength of railing anchor bolts is usually controlled by anchorage failure. Impact makesbad anchorage conditions worse. Designers should consider the findings of three otherreports15 and strive for the most ductile connection possible under the conditions imposed.



Typical anchorage details currently in use are shown in Figure 9-41, Figure 9-42, andFigure 9-43.

Figure 9-41. : Example of Sign Bridge Anchorage (Online users can click here to view thisillustration in PDF.)



Figure 9-42. : Example of High Mast Illumination Pole Anchorage (Online users can clickhere to view this illustration in PDF.)



Figure 9-43. : Example of Integral Steel Bent Cap Anchorage (Online users can click hereto view this illustration in PDF.)

Chapter 9 — Special Designs Section 13 — Deck Drainage


Section 13 Deck Drainage

Background

Most bridge engineers agree that some provision must be made to get rainwater off thebridge. Hydraulic analysis has probably been avoided until recently, but some drainageprovisions have always been shown on the details.

At one time, no bridge was allowed to be constructed to a flat gradeline because of drainageconsiderations. This is still a good idea.

Roadway crowns started as straight slopes either way from the center, went briefly toparabolic crowns, then simulated parabolas and, finally, back to straight slopes either wayfrom the center. Cross-slope magnitude has varied considerably, but 0.02 ft. per ft., in atypical tangent section of roadway, is the current preference.

Since most bridges had curbs originally and now have concrete wall railing, some type ofdeck drain has usually been provided. One of the earliest was 2 in. round holes at 4 ft.spacing. A sheet metal drain box, open at the lower curb face, was used for several years onconcrete girders. Cast iron soil pipe was used on many steel beam and concrete girderbridges. The number of drains stabilized at two per span or two per concrete placement oncontinuous units with drains to be omitted over header banks. By the early 1940s theaccepted drain detail became a 4 by 6 in. formed opening in the deck slab. Drip beads, in theform of 3/4 in. triangular depressions, were provided around each drain outlet. These werespaced approximately 20 ft. apart with a note to “omit over header banks and railroadtracks.” This drain is still used, but in the 1960s, bridge designers abandoned any claim tohydraulic expertise with a plan note that drains shall be “spaced as directed by theEngineer.”

For awhile, many bridges were constructed with no curbs and open railing. Continuous dripbeads were provided beneath the overhanging slab. This eliminated the drainage problem,except in de-icing salt areas where small curbs were added to limit corrosive runoff.

Hydroplaning became an issue affecting bridge decks as well as approach pavement.Texturing concrete surfaces to an average depth of 0.05 in. was the solution. This providessufficient water storage to reduce the hydroplaning film to acceptable thickness.

There is an increasing demand for closed drainage systems for bridges, because ofaesthetics, protection of traffic below, or water quality issues.

Design was by intuition for most open drainage systems. Closed systems usually requiredthe services of a hydraulic engineer. FHWA publications in 198416 and 198617 containvaluable guidance for the design of bridge deck drainage.

Recommended drain design has changed significantly since 1988.



Maintenance problems can be acute. Open drains usually let runoff get to the superstructure,causing corrosion in de-icing areas. Improper provision for drainage of the approachroadway has caused erosion of the embankment. Grate inlets used for closed systems arehighly susceptible to clogging by roadway debris.

Aesthetics can be compromised by either type of system. Staining from open drainage iscommon. Closed systems require downspouts, which are often handled in an unsightlymanner.

Current Status

Bridge deck drainage is the responsibility of district highway engineers, assisted by theHydraulics Section and the Bridge Design Section.


The Hydraulics Section should be consulted if there is any doubt as to the deck drainagerequirements or if a closed system must be designed.

A few suggestions are offered, as follows:

♦ For various design parameters and coefficients see the Hydraulic Design Manual.

♦ Open or slotted railing is generally recommended for stream crossings except in the de-icing areas, or areas with water quality issues.

♦ Do not use deck drains unless absolutely necessary.

Note: One report18 offers a quick method for determining the maximum undrainedbridge length. Be sure that the approach facilities are capable of handling the bridgerunoff.

Caution: Beware of cross-slope reversals, which may direct the entire gutter flow acrossthe roadway.

♦ Open deck drains located “as directed by the Engineer” should be avoided.

♦ In de-icing areas open deck drains should empty below the superstructure as shown onbridge standards IBMS, SBMS, and CG-MD.

♦ Closed deck drainage systems should have grate inlets and PVC downspouts encased inthe substructure insofar as practical.

Note: Discharge may be into a storm sewer, stilling well, or close to the ground if waterquality issues are not a major concern. A cleanout plug should be provided.

Note: TxDOT standards include a guidance detail sheet Bridge Drain Details thatshows the latest recommended cast steel grate inlet for use on bridges with conventionaldecks. Instructions on the sheet require the designer to modify the sheet for specificapplications. This sheet is not to be used as a standard.



Examples: A welded steel grate inlet system and drainage installation used onsegmental box girder bridges are shown in Figure 9-44 and Figure 9-45, respectively.

Requirement: If multiple drains are required in sag areas, structural integrity of thebridge deck should be maintained by 3 ft. minimum drain spacing.

Figure 9-44. : Example of Grate Inlet and Installation for a Segmental Superstructure(Online users can click here to view this illustration in PDF.)



Figure 9-45. : Example of Drainage Installation for a Segmental Superstructure (Onlineusers can click here to view this illustration in PDF.)

Chapter 9 — Special Designs Section 14 — Reinforced Concrete Box Culverts


Section 14 Reinforced Concrete Box Culverts

Background

Texas standard box culvert detail sheets can be found dating from 1918. Types of culvertsidentified are:

♦ Laminated timber

♦ Patented creosoted timber

♦ Stone walls with stone slab

♦ Stone walls with precast reinforced concrete slab

♦ Vitrified clay segmental block, round, or flat bottom arch

♦ Masonry arch

♦ Interlocking precast concrete u-shaped sections

♦ Concrete wall with footing and reinforced concrete simple slab

♦ Cast-in-place single boxes, reinforced for positive moment only

♦ Cast-in-place single boxes, reinforced as a frame

♦ Cast-in-place multiple boxes, reinforced as a frame

♦ Precast single box sections

♦ Precast two-piece single box sections

The most widely used culverts are the reinforced concrete single and multiple boxes.

Early Designs

From the late 1940s to the middle 1990s, TxDOT maintained an extensive set of culvert andwing wall standard detail sheets. Used directly, or modified and used for higher fillinstallation, they were the basis for many millions of dollars worth of culvert construction.

Design procedures for these boxes did not exactly conform to AASHTO requirements. Theoriginal 1948 standard designs used service load methods and were based on theseassumptions:

♦ Vertical earth pressure 120 pcf times 0.7

♦ Lateral earth pressure 30 pcf equivalent fluid

♦ Live load was a 12 kip wheel

♦ Live load distribution according to a Westergaard article in “Public Roads,” March1930

♦ Two feet of surcharge and full lateral pressure used for corner moments



♦ No lateral pressure used for positive moments

♦ Live load in one span only for positive moment

♦ Allowable stresses in concrete, 1,000 psi and steel, 18,000 psi

♦ Shear in slabs not considered

♦ Moment distribution by hand calculations

As changes were made to the AASHTO Specifications and local practice, the details weremodified but, because of the large number of design combinations and time constraints, themodifications were not exactly accurate for all sections. This makes it difficult to verifythese designs according to strict AASHTO requirements. Culverts constructed by thesedesigns have proven adequate by virtue of their performance under traffic.

In spite of their somewhat antiquated design, no significant malfunctions have beenobserved in the many culverts constructed to these details. This was one reason for thereluctance by TxDOT to redesign according to the latest specifications.

Significant to advancement of culvert technology was acceptance of the precast box culvert.For Texas the industry began in Beaumont in the early 1970s with a fabricator makingstandard cast-in-place sections vertically in 8 ft. lengths. High-strength dry concrete wasrequired because the inside form was a mandrel that was retracted as soon as concreteplacement was complete. Outside forms were also removed soon.

Another fabricator, in Harlingen, also provided TxDOT with standard precast box culvertdetails before acceptance of an industry standard covered by ASTM C789.

AASHTO acceptance was slow because ASTM C789, rather than the usual materialspecification, actually gave the live load and fill height that each section could sustain. Thisusurpation of the bridge engineer’s field of authority plus loading research in progress andquestionable shear transfer across the joint in direct traffic situations, made it difficult for theindustry to gain formal acceptance from AASHTO. Texas allowed the industry standardsection prior to acceptance, but established allowable fill heights according to local practice.

ASTM C850 was published to cover direct traffic precast box culverts. Several TxDOTstandard detail sheets were developed to allow the use of precast products.

Recent Changes

In the early 1980s, the specifications allowed culverts to be bid by the linear foot, instead ofcubic yard. This allowed the contractor to decide whether to build precast or cast-in-placeculverts. Virtually all culverts are now bid this way, as shown in the "Reinforced ConcreteBox Culvert Usage" table.

Another advance in culvert design began in the early 1980s when the Federal HighwayAdministration insisted action be taken to protect errant motorists from plunging into thespace at the end of cross drainage culverts or running into the headwalls of parallel drainageculverts. The solution involved installing pipe runners across the opening, which createdhydraulic concerns because of negative effect on the ability of the culvert to carry storm



water. Research was conducted at CTR (Report 301-1F) on the hydraulic aspects and at TTI(Report 280-1 and 280-2F) on the structural aspects. The structural results were fine-tunedand safety end treatment standard detail sheets were prepared.

In the late 1990s it became necessary to produce box culvert standard detail sheets in metricunits, and the entire culvert series was redesigned, using load factor design and the currentspecifications. These metric box culvert standard detail sheets were then converted toEnglish units in 2000. At that time significant changes were made to the wing wall andsafety end treatment details.

Current Status

The current TxDOT box culvert standard detail sheets can account for almost any boxculvert design need. Culvert standard detail sheets are available on the TxDOT web site forthe following box culverts and appurtenances:

♦ Cast-in-place single boxes (30 ft. max. fill height)

♦ Cast-in-place multiple boxes (23 ft. max. fill height)

♦ Precast single boxes (12 ft. to 20 ft. max. fill height depending on span)

♦ Wing walls (straight, flared, and parallel)

♦ Safety end treatments


The TxDOT box culvert standard detail sheets will significantly reduce the need for specialdesigns for culverts and end treatments. However, if a special design is warranted, somedesign parameters are as follows:

♦ Vertical earth pressure 120 pcf

♦ Lateral earth pressure 40 pcf

♦ Live load is a 16 kip wheel with impact per AASHTO

♦ Distribution of a wheel is a square of 1.7 x fill depth, for fills ≥ 2 ft.

♦ For fill heights less than 2 ft., load is considered a point load. Distribution and designper AASHTO slab design requirements.

♦ Two feet of surcharge and full lateral pressure used for corner moments

♦ Half lateral pressure used for positive moments

♦ Spans loaded according to influence lines for moments

♦ Class C concrete, f 'c = 3,600 psi; Grade 60 reinforcing steel, fy = 60,000 psi

♦ Slab thickness based on allowable shear



Additional information, including when and where to use culverts, can be found in theRoadway Design Manual. A discussion on the hydraulic requirements of culverts can befound in the Hydraulic Design Manual.

Reinforced Concrete Box Culvert UsageContract Bid Quantities

Calendar Concrete for Culverts (C.Y.) 4 Box CulvertYear 1 Class A 2 Class C 3 Class S (L.F.)1966 148,000 10,000 1971 730 130,000 1976 39,000 3,000 1981 23,000 2,000 1983 7,500 2,000 100 49,0001986 19,000 300 250 246,7001988 11,000 2,000 100 226,7001997 210 548 56 187,612

1. Five sack 3,000 psi2. Six sack 3,600 psi3. Special deck concrete for direct traffic culverts4. Alternate precast or cast-in-place; linear feet of single barrel

Chapter 9 — Special Designs Section 15 — Reinforced Concrete Pipe


Section 15 Reinforced Concrete Pipe

Background

Specifying reinforced concrete pipe for use in projects was somewhat disorganized. In theTexas Standard Specifications (1982), reinforced concrete pipe could be bid as “culvert” or“sewer.” Bedding for both was specified under the “culvert” item. Excavation for culvertswas specified in the Structural Excavation item. Excavation for sewers was specified inmore detail in the “Excavation and Backfill for Sewers” item. Monolithic concrete pipe andcorrugated metal pipe could also be bid under the “sewer” specification. Moreover, theimplication that open-ended cross drainage structures are “culverts” and partially closeddrainage systems with inlets and manholes are “sewers” was not always followed.

To do away with some of the confusion with specifying reinforced concrete pipe for aproject, changes to specification concerning reinforced concrete pipe were included in TexasStandard Specifications (1993). The change in specifications allowed all reinforcedconcrete pipe, used as culvert or sewer, to be specified under one item number. Excavation,bedding, and backfill, regardless of the application, are specified under one item number aswell.

Current Status

As indicated on the "Reinforced Concrete Pipe Usage" table reinforced concrete pipe is usedextensively in Texas highway construction. Virtually all of it is round pipe and most is from12 to 48 in. diameter.

There are four common types of concrete pipe installation conditions:

♦ Trench

♦ Positive projecting

♦ Negative projecting

♦ Imperfect trench

Descriptions and illustrations of these installation conditions, as well as recognized classesof bedding can be found in the Hydraulic Design Manual.

Appurtenances are many and vary in detail. The current TxDOT culvert standard detailsheets can account for many these needs. Culvert standard detail sheets are available on theTxDOT web site for the following applications related to reinforced concrete pipe:

♦ Headwalls with flared or parallel wings

♦ Manholes

♦ Inlets

♦ Slotted drains



♦ Safety end treatments

Pipe strength may be specified as one of the classes described in ASTM C76 or as “D-Load.” When specifying pipe some guidelines are as follows:

♦ When pipe is specified using ASTM C76 class, the location of each class must beshown on the plans.

♦ When pipe is specified using one ASTM C76 class for a project, a brief note on theplans verifying the class required will be sufficient. Class III is the most often used.

♦ When pipe is specified using D-Loads, values of the D-Load for the various runs of pipemust be given in the plans.

♦ Structural design consideration is required to insure adequate strength for theinstallation conditions, type of bedding, and fill height anticipated.

Specification by D-Loads required by ASTM C76 is acceptable. Specification by ASTMClass in the bid item is also acceptable if the advantages of manufacturer design appearremote.

Specification of concrete pipe by D-Load allows the manufacturer to determine the mosteconomical combination of wall thickness and reinforcing steel. The smaller the incrementalmagnitude of specified D-Loads, greater will be the problem of pipe location on the project.

ASTM C76 classes, and equivalent recommended minimum D-Load increments for eachclass, can be found in the Hydraulic Design Manual.

Headwalls and wing walls must be Class C concrete as required by Texas StandardSpecifications (1993), Item 466 “Headwalls and Wing walls” and the current SpecialProvisions thereto.



Reinforced Concrete Pipe UsageQuantity Let to Contract

12 Months Ending November 1998Round Pipe Diameter

(in.)Pipe Culvert

(L.F.)Pipe Sewer

(L.F.)12 16,928 15 23 2018 54,66421 24 7,052 41,06827 39 30 3,723 28,68333 86636 2,171 37,24839 42 2,660 25,89948 3,615 9,29654 525 8,03360 5,17366 1,34272 8,70578 65684 88 1,296

♦Total 36,736 222,949Arch 1,806

Elliptical 40♦ D-Load specified for all pipe culverts, ASTM C76 class specified for 94

percent of the remainder.


The design of reinforced concrete pipe shall be governed by the AASHTO Specifications.Some design notes are as follows:

♦ Vertical loads are calculated with empirical equations developed from analytical andexperimental observations of soil-structure interaction.

♦ Earth loads vary with properties of the fill, subgrade, pipe, and type of installation.

♦ The relationship of actual support conditions to the three-point bearing test isestablished by empirical equations based on the class of bedding. Copious table andcharts are provided to assist in this.

♦ Live load is calculated according to AASHTO distribution of an HS20 truck.



Additional design guidance can be found in the American Concrete Pipe Association(ACPA) Concrete Pipe Design Manual.

D-Load required is an expression of the calculations explained above in pounds per foot oflength per foot of inside pipe diameter. D-Load provided is the load at which 0.01 in. crackforms in a representative section of pipe under the three-point bearing test described inASTM C497. Use of pipe certified for a D-Load equal to or greater than required by theplans or ASTM C76 insures serviceability of the pipe.

Alternatively, direct design of concrete and reinforcing in the pipe may be performed withthe aid of a finite element soil-structure interaction program. Use of this method by TxDOTis unlikely.

Required D-Loads greater than 3000D should be submitted to the TxDOT Bridge Divisionfor special design. Variations of trench width, backfill type, or bedding may be moreeconomical than the increase in pipe strength. The AASHTO Specifications and the ACPAConcrete Pipe Manual will guide such designs. Additional industry guidance may also besolicited.

D-Loads required for various fill heights and diameters of pipe, based on installation andbedding conditions, can be found in the Hydraulic Design Manual.

Allowable fill heights for concrete arch pipe and horizontal elliptical pipe can be found inthe Texas Standard Specifications (1993).

Hydraulic design criteria can be found in the Hydraulic Design Manual. Conduit durabilitydesign criteria can be found in the Hydraulic Design Manual.

Chapter 9 — Special Designs Section 16 — Corrugated Metal Pipe


Section 16 Corrugated Metal Pipe

Current Status

Corrugated metal pipe usage is shown in the "Corrugated Metal Pipe Usage" table.Virtually all is steel pipe, and 67 percent is pipe arch.

Appurtenances are many and vary in detail. The current TxDOT culvert standard detailsheets can account for many of these needs. Culvert standard detail sheets are available forthe following applications related to corrugated metal pipe:

♦ Headwalls

♦ Manholes

♦ Inlets

♦ Slotted drains

♦ Safety end treatment

Pipe strength is specified as a wall thickness for a given size and maximum fill height.Bedding and backfill are critical.

Steel and aluminum pipe are available in several different corrugation configurations,aligned either annularly or helically with riveted, resistance welded, or mechanically lockedseams.


The design of corrugated metal pipe shall be governed by the AASHTO Specifications.

Additional design guidance can be found in the American Iron and Steel Institute (AISI)book Modern Sewer Design.

Allowable fill heights and wall thicknesses for full circle corrugated metal pipe can be foundin the Hydraulic Design Manual.

Allowable fill heights and wall thicknesses for steel and aluminum pipe arches can be foundin the Texas Standard Specifications (1993) Item 460 “Corrugated Metal Pipe.”

Hydraulic design criteria can be found in the Hydraulic Design Manual. Conduit durabilitydesign criteria can be found in the Hydraulic Design Manual.

Chapter 9 — Special Designs Section 16 — Corrugated Metal Pipe


Corrugated Metal Pipe UsageQuantity Let to Contract

12 Months Ending November 1998Round Pipe Pipe Arch

Diameter(in.)

Bid Quantity(L.F.)

DesignSize

Bid Quantity(L.F.)

6 8

12 640 15 473 1 1218 5,559 2 8,54921 36 24 5,125 3 26,04830 1,653 4 4,39936 1,078 5 1,76642 984 6 1,93848 482 7 79754 8 73560 95 9 66 10 72 201 11 78 12 84 13 90 886 14

102 126 19 226

♦ Total 17,212 44,470♦ Only 0.3 percent had bituminous coating. Only 48 linear feet had paved invert.

Chapter 9 — Special Designs Section 17 — Structural Plate Structures


Section 17 Structural Plate Structures

Current Status

Structural plate structures include the following:

♦ Structural plate pipe

♦ Pipe arches

♦ Underpasses

♦ Box culverts

♦ Special shapes

They may be aluminum or steel and may or may not be elongated prior to backfill. Thestructures may be partially prefabricated or field fabricated.

The usage of structural plate structures by TxDOT is very small.

There are no standard detail sheets available from the Bridge Division for end treatment orfootings, nor are there any structural design guides.

Galvanized steel plate is available with 6 in. x 2 in. annular corrugations. Aluminum platehas 9 in. x 2 1/2 in. corrugations. Seams are bolted.


Designs are usually initiated by industry and checked by the TxDOT Bridge Division, priorto construction, using the AASHTO Specification and American Iron and Steel Institutebook Modern Sewer Design as guides.

Chapter 9 — Special Designs Section 18 — Long Span Structural Plate Structures


Section 18 Long Span Structural Plate Structures

Background

Long span structural plate structures were constructed with the same corrugated plate usedfor structural plate structures, but they required special stiffening attachments to limitdeflection and stress. They also required careful backfilling to prevent distortion of theshape. Several different shapes were available, such as horizontal ellipse, low profile arch,high profile arch, inverted pear, and pipe arch.

In the early 1970s, various manufacturers of corrugated metal plate actively marketed longspan structures in Texas. Designs were highly proprietary.

Several installations were made. All were very critical in the backfilling stage. One, inparticular, incurred so many problems as to discourage most districts from further use. Thestructure was a horizontal ellipse with 40 ft. span. It was completed after much delay andadditional strengthening but retained some abnormal deformation of the top. Recentlyadditional deformations have appeared in the bottom. Cause of the malfunction had to berelated to backfill sequence and/or equipment or design. Whatever the cause, thisexperience, and a few others nationwide, established the delicate reputation of long spanplate structures.

The 1982 TxDOT Standard Specifications covered long span structural plate structures.Only three such structures were constructed on the state highway system through 1997.

1987 High profile arch (aluminum) 292 x 178 in.

1986 Low profile arch (steel) 415 x 136 in.

1984 Low profile arch (aluminum) 364 x 119 in.

This item was dropped from later construction specifications.

Current Status

The only long span structural plate structure in 1998 was 240 x 120 in. let under the Item461 Structural Plate Structures of the 1993 TxDOT Standard Specifications.


If there should be a need for a long span structural plate structure, the design will probablybe initiated by industry and should be checked by the TxDOT Bridge Division using theAASHTO Specification as a guide.

Chapter 9 — Special Designs Section 19 — Sign Support Structures


Section 19 Sign Support Structures

Background

Construction on the interstate highway system demonstrated a need for prominent trafficsigns. Signs alongside the freeway were often inadequate.

Regulatory and precautionary signs could remain fairly small. Some directional signs, ifsufficiently large, could be mounted alongside the freeway. Many conditions requireddirectional signs to be mounted above the freeway, or ramp lanes, to be effective.

The first Bridge Design Section standard details of sign supports for interstate highwaysbegan to be issued in 1960. The first version made extensive use of galvanized pipe, evenfor overhead truss bridges. This proved to be expensive. Some of the larger districts,especially Houston Urban Project, designed their own sign supports. Commerciallydesigned options were also allowed.

The standards were revised in 1963 to replace much of the galvanized pipe with wide flangeand angle shapes. Some districts and fabricators still dislike them, and the practice ofallowing commercial design options continued. Many different concepts emerged from this.

Early roadside sign mounts created alarm from a traffic safety standpoint because of theirexposed drilled shaft foundations and strong wide flange supports. Responding to pressurefrom highway engineers, one person in the Bridge Design Section coordinated a largeresearch study19 to develop breakaway supports for sign support structures. Mechanismswere developed for small and large roadside signs that have saved many lives. Oneoverhead sign bridge was tested with breakaway supports, which turned out to be feasiblebut impractical. The bridge is now in service near Hearne.

In 1966 responsibility for sign support design was transferred to the Traffic Operationssection of the Maintenance and Operations Division. Roadside sign supports wereconstructed from Bridge Design Section standard details, but overhead sign bridgescontinued to be constructed mostly according to commercial designs.

The AASHTO Specifications for Structural Supports for Highway Signs, Luminaires, andTraffic Signals were introduced in 1968, but were heavily revised to the present condition in1975. At about the same time, it was decided that the Bridge Design Section shouldreassume responsibility for overhead sign bridge design. There ensued a hectic periodduring which new and different specification controls were enforced on fabricators longaccustomed to following more liberal rules for optional designs. Simultaneously, newstandard details were being developed. With input from one such fabricator, the newspecification controls were interpreted in the most economical manner possible. Optionaldesigns began to disappear and in 1984, specifications were revised to require only thestandard details.

Overhead sign support usage for the period 1980-1989 is shown on Figure 9-46.



Figure 9-46. : Chart of Highway Accessory Structures Usage (Online users can click here toview this illustration in PDF.)

Design Issues

Design of overhead sign bridges, begun with little guidance, was recognized by AASHTO in1968 and underwent a complete revision in 1975. The controlling load for sign bridges iswind. For cantilevers, it is eccentric dead load in combination with wind. Forces from a 50-year recurrence interval wind are resisted at 1.4 times the yield strength of the material.There is a feeling that this is conservative, since no sign bridges have failed due to wind, noteven the light commercial designs.



Construction Issues

Construction problems centered on quality of galvanizing in the beginning. A few years ofinteraction between TxDOT and the hot dip galvanizing industry transpired before detailsand capabilities were optimized and the appearance was satisfactory with constructionengineers. Coordination of foundation spacing and elevation with fabricated dimensions ofsign bridge components can be a problem. Conditions often change during construction andmay not be recognized until late in the project. There is very little provision for adjustmenton a sign bridge. Getting the right anchor bolts installed to the proper dimensions on thefoundation has also been prone to error.

Maintenance Issues

Maintenance problems have been caused by galvanizing deficiencies. A number of tubularsupport members, mostly for cantilever or butterfly sign bridges, have rusted severely fromthe inside because of incomplete zinc coverage. The details provided insufficient openingsto allow a free flow of zinc within the tubes, resulting in certain areas being coated withpickling acid rather than zinc.

Current Status

Supports for signs are a responsibility of the Maintenance and Operations Division.

All overhead sign support structures are constructed from the standard details prepared bythe Bridge Design Section.


Standard overhead sign support structure designs, and special designs that may be required,will conform to the AASHTO Specification for Structural Supports for Highway Signs,Luminaires, and Traffic Signals, with a few clarifying local interpretations. Foundations arealways drilled shafts, designed by Bridge Design Section methods, and verified byresearch.20 Foundation design curves are included with the standards. There should be asoil test boring no further than 500 ft. from each foundation for proper design.

Overhead sign support structures should not be mounted on bridges unless absolutelynecessary. If located within the limits of a bridge, support brackets are usually preferable toground mounting. Support bracket of a type previously used for the smaller sign bridgecolumns is shown on Figure 9-47.

Questions regarding overhead sign supports or their mounting should be directed to theBridge Design Section.

Drilled shaft foundation design curves for roadside sign supports, developed on a 1970research project,21 are included with the standard details.



Figure 9-47. : Example of Overhead Sign Bridge Support Bracket (Online users can clickhere to view this illustration in PDF.)

Chapter 9 — Special Designs Section 20 — High Mast Illumination Poles


Section 20 High Mast Illumination Poles

Background

These structures have been exclusively steel poles since the concept of high level lightingbegan in Texas in the early 1970s. The mounting height was originally 150 ft. Designsoriginated with a manufacturer and were checked structurally by the Bridge Design Section.Soon several fabricators were pursuing the market and it was necessary to fine-tune designprocedures in order to referee impartially between them.

Poles were 8-sided, 12-sided, or round. They were tapered in diameter and spliced byinserting the top of one section into the bottom of the next with a 1.5 diameter overlap.Most of the Texas installations were galvanized. Most had single drilled shaft foundations.

A circular luminaire mounting ring was connected to a retainer on top of the pole. Gears,pulleys, and cables were located inside the pole for raising and lowering of the luminaires.Mechanical features and lights were the responsibility of the Highway Design Section. TheBridge Design Section was responsible for structural design of the pole.

High-level illumination pole usage for the period 1980-1989 is shown on Figure 9-46.

Design conformed to the 1968 AASHTO Specification for 50 year recurrence interval windspeed. Foundation design was a compromise between a Broms method22 and U.S.Agriculture Department recommendations for pole buildings.

Design competition was quickly manifest in the Houston Urban Expressway project and, forseveral years, designs from both offices were constructed. Finally, with completion of theBridge Design Section proposed standard details in 1989, agreement was reached wherebyboth offices would use the same design for high mast poles.

Scratch gauges were mounted on a pole in three different locations in an attempt to evaluatefatigue stresses near the base. The only measured stresses greater than 5 ksi were a fewexcursions during a norther near Canyon.

Minor construction problems were encountered with shop welding, fit of the lap splices,type of anchor bolt, and proper placement of lights.

The poles have been practically maintenance-free. The only reported failure was collapse ofthe upper section of a pole in a tornado near Wichita Falls. There was some crackingreported in the pole-to-base plate weld on some early installations. This was determined tobe cracks in the zinc coating. Other states have reported failure of lap joints due to rustbuildup in weathering steel poles. Texas has only one such installation with no evidence ofmalfunction.

Chapter 9 — Special Designs Section 20 — High Mast Illumination Poles


Current Status

Bridge Division standard drawings HMIP 80, HMIP 100, and HMIF, dated November 1986and standard construction specification Item 613 are used for high mast illumination poles.The Houston District uses a modified version of these details.

New standard drawings HMIP and HMIF, dated May 1989, are available but are awaiting anew construction specification before formal use. The details cover 8- and 12-sided poles of100, 125, 150 and 175 ft. heights designed for 100 mph wind. The constructionspecifications will allow optional designs and require galvanized poles.


Standard designs comply with the AASHTO Specifications. Linear beam column programBMCOL51 is useful for calculating moments that include the magnification effect of axialload and deflection. Tip deflection at design wind is upward of 10 ft.

Foundation design curves, shown on the standard drawings, were developed using nonlinearBMCOL7623 with lateral soil resistance based on various research reports. Soil was allowedto reach half ultimate stress under the action of design wind.

Chapter 9 — Special Designs Section 21 — Traffic Signal Poles


Section 21 Traffic Signal Poles

Background

Traffic signal poles present a complicated design problem that was oversimplified by earlyconstruction specifications.

They may be classified as follows:

♦ Pedestal poles: Signal mounted directly atop the pole

♦ Strain poles: Signals suspended from cables strung between poles

♦ Mast arm poles: Signals mounted at the ends of horizontal cantilever arms. Arms maybe attached to one or two sides of the pole and can extend as much as 40 ft.

♦ Bridge support poles: Signals suspended from a beam supported by two poles. Note thatthis is seldom used in Texas.

Poles are erected under the bid item, “Installation of Highway Traffic Signals,” by the lumpsum for each project. Each project may contain one or more poles, steel or timber, furnishedby the contractor or by TxDOT. It is practically impossible to determine the number oftraffic signal poles used by the department. The number of signal projects for the period1980-1989 is shown on Figure 9-46.

Many poles have been purchased by the state for stock, some of which are used in district-constructed signal installations and others furnished to contractors for projects involvingsignals only.

The Bridge Design Section assumed responsibility for traffic signal pole design in 1975along with overhead sign supports, and utter confusion ensued. Poles were furnished to aprocurement or construction specification that prescribed a design load and anchor bolt sizebut required a design check based on the AASHTO Specification. During shop plan review,the specified design load was sometimes found to be unrepresentative of the actual loadingconditions. Often, the specified anchor bolts did not conform to the actual loads and currentdesign practice. Further delay was caused by shop plan submission procedures. Someprojects were held up, and others had anchor bolts installed before approved shop plans,requiring different anchor bolts, were received. The situation was gradually improvedthrough interaction between the Bridge Design Section, Traffic Operations Section, districts,contractors, and fabricators.

The Bridge Design Section prepared, and the Traffic Operations Section issued, ninestandard drawings in June 1985, covering strain pole and mast arm pole assemblies. If thedistricts continue to use these drawings, problems will be minimized.

Chapter 9 — Special Designs Section 21 — Traffic Signal Poles


Current Status

If standards SP-80, SP-100, SMA-80, SMA-100, DMA-80, DMA-100, MA-C, MA-D, andTS-FD are used for the conditions shown on the drawing, submission of shop plans to theBridge Design Section is not required.


For special designs, shop drawing submittal is required, and designs will be checkedaccording to the AASHTO Specification for Structural Supports for Signs, Luminaries, andTraffic Signals. For design purposes, the yield stress of anchor bolts will be assumed to beno more than 55 ksi and, for pole material, no more than 50 ksi.

Chapter 9 — Special Designs Section 22 — Sound Walls


Section 22 Sound Walls

Background

When the U.S. 281 freeway in San Antonio was designed in the early 1970s, extensive andarchitecturally pleasing walls were required to insulate some sensitive areas from highwaynoise. Since then there has been a growing demand for such treatment.

Federal requirements for sound abatement have been promulgated. District environmentalcoordinators ensure response to these requirements on each project.

Lately, a majority of sound walls have been constructed in the Houston District, where theyhave developed a design system and construction specification. The Bridge Design Sectionhas assisted other districts, sporadically, in the structural aspects of sound walls. Drilledshafts have been the preferred foundation for ground-mounted walls.

Construction specifications usually allow commercial alternates for the appearance portion.Commercial alternates have predominated, with some sort of masonry or precast concretepanels being the predominant types.

There have been no maintenance problems to date, except for one instance where a streetsweeper knocked a portion of wall into an empty playground. Concern for this problem hasled Houston to require extra strength in the lower part of walls.

Sound barriers mounted atop bridge rails must be able to resist vehicle impact. Walls havebeen mounted behind bridge rails to ease this concern; however, a newly designedcombination T501 shape and 7 1/2 in. thick reinforced concrete wall system has recentlybeen successfully crash tested.

Current Status

The Bridge Design Section will be glad to assist any district in the structural aspects ofsound wall design.


The AASHTO Guide Specification for Structural Design of Sound Barriers should befollowed.

A good reference for masonry walls is the TEK Manual published by the National ConcreteMasonry Association.

Foundation design should be referred to the Geotechnical Group.

Chapter 9 — Special Designs Section 23 — Wildlife Issues


Section 23 Wildlife Issues

Background

With the increase in environmental concerns has come an awareness of the need to considerthe impact of highway construction on the flora and fauna of Texas.

TxDOT has a long history of wildflower cultivation in the right-of-way. Grasses, plants,and decorative treatments have increasingly been used to make the highwaysenvironmentally friendly. Small pipe underpasses were constructed near Bastrop to let theendangered Houston Toad cross safely.

In the mid-1980s, the Congress Avenue Bridge in downtown Austin became home to one ofthe largest urban bat colonies in the world, providing roosting space for 1 1/2 millionMexican freetail bats. Since then, efforts have been made to purposefully provide bathabitat in other bridges and culverts in Texas and throughout the U.S.

Cliff swallows have nested under bridges near water in rural and suburban settings for a longtime. For obsolete bridges that are to be replaced, it may be advisable to provide birdnetting to exclude swallows approximately one year prior to demolition. Before disturbingnesting swallow colonies, determine whether a permit is required. Netting details andsuggestions are available from the Bridge Division.

Near cities, pigeons perch on concrete and steel bridges alike. Pigeon nests and droppingsare reported to be a source of corrosion to steel bridge members. Large pigeon populationsroosting in bridges over rivers also cause concerns with water quality during low flowperiods.

Current Status

The Bridge Design Section will work with the Environmental Division to mitigate theeffects of bridges on Texas wildlife. The Special Projects Branch is the designated contactfor wildlife matters.


There is a bat and bridges project report issued by Bat Conservation International(www.batcon.org) that contains suggestions for providing bat habitat. Pigeon excludershave been developed for use in critical areas.

1 “Iron and Steel Beams,” 1883-1952, American Institute of Steel Construction, Fifth Printing 1968.2 “Guide Specification for Strength Design of Trusses,” AASHTO, 1985.3 “Arch Bridges,” Structural Engineering Series No. 2, FHWA, 1977.4 “Construction and Design of Cable-Stayed Bridges,” Podolny, W. Jr. and J.B. Scalzi, John Wiley and Sons, 1976.5 “Recommendations for Stay Cable Design and Testing,” PTI Ad Hoc Committee on Cable Stayed Bridges, 1986.6 “Design of Neoprene Bridge Bearing Pads,” E.I. duPont de Nemours & Co., 1959.7 “Elastomeric Bearing Research, NCHRP Report 109, 1970.

Chapter 9 — Special Designs Section 23 — Wildlife Issues


8 “Elastomeric Bearings Design, Construction and Materials,” Stanton, J.F. and C.W. Roeder, NCHRP Report 248,1982.9 “Performance of Elastomeric Bearings,” Roeder, C.W. and others, NCHRP Report 298, 1987.10 “Engineering Properties of Neoprene Bridge Bearings,” DuPont Company, 1983 Est.11 “Elastomeric Bearing Research, NCHRP Report 109, 1970.“Engineering Properties of Neoprene Bridge Bearings,” DuPont Company, 1983 Est.12 “Development Length for Anchor Bolts,” Breen, J.E., CFHR, Report 53-1F, 1964. “Factors Affecting Anchor Bolt Development,” Lee, D.W. and J.E. Breen, CHFR, Report 88-1F, 1966.“Strength and Behavior of Anchor Bolts Embedded Near Edges of Concrete Piers,” Hasselwander, G.B and others,CFHR, Final Report 29-2F, 1977.“Axial Tension Fatigue Strength of Anchor Bolts,” Fischer, F.L. and K.H. Frank, CFHR, Report 172-1, 1977.“Fatigue of Anchor Bolts,” Frank, K.H., CFHR, Final Report 172-2F, 1978.“Strength and Behavior of Bolt Installations Anchored in Concrete Piers,” Jirsa, J.O. and others, CTR, Final Report305-1F, 1984.“Response of Highway Barriers to Repeated Impact Loading,” Klingner, R.E. and others, CTR, Report 382-1 & 2F,1985.“Design Guide for Steel to Concrete Connections,” Cook, R.A. and others, CTR, Report 1126-4F, 1990.“Fatigue Loading in Sign Structures,” Creamer, B.H. and others, CFHR, Report 209-1F, 1979.13 “Development Length for Anchor Bolts,” Breen, J.E., CFHR, Report 53-1F, 1964.14 Development Length for Anchor Bolts,” Breen, J.E., CFHR, Report 53-1F, 1964. “Factors Affecting Anchor Bolt Development,” Lee, D.W. and J.E. Breen, CHFR, Report 88-1F, 1966.“Strength and Behavior of Anchor Bolts Embedded Near Edges of Concrete Piers,” Hasselwander, G.B and others,CFHR, Final Report 29-2F, 1977“Strength and Behavior of Bolt Installations Anchored in Concrete Piers,” Jirsa, J.O. and others, CTR, Final Report305-1F, 1984.15 “Bridge Deck Design for Railing Impact,” Arnold A. and T.J. Hirsch, TTI, Final Report 295-1F, 1983.“Response of Highway Barriers to Repeated Impact Loading,” Klingner, R.E. and others, CTR, Report 382-1 & 2F,1985.“Design Guide for Steel to Concrete Connections,” Cook, R.A. and others, CTR, Report 1126-4F, 1990.16 “Drainage of Highway Pavements,” Hydraulic Engineering Circular No. 12, FHWA, TS-84-202, 1984.17 “Bridge Deck Drainage Guidelines,” Turner-Fairbank Highway Research Center, Report No. FHWA/RD-87-014,1986.18 “Bridge Deck Drainage Guidelines,” Turner-Fairbank Highway Research Center, Report No. FHWA/RD-87-014,1986.19 “Modern Sewer Design,” American Iron and Steel Institute, First Edition, 1980.20 “Analysis of Drilled Shaft Foundation for Overhead Sign Structures,” Meyer, B.J. and L.C. Reese, CFHR, FinalReport 244-2F, 1980.21 “Design Procedure Compared to Full Scale Tests of Drilled Shaft Footings,” Ivey, D.L. and W.A. Dunlap, TTI,Report 105-3, 1970.22 “Tapered Steel Poles – Caisson Foundation Design,” Teng and Associates, U.S. Steel Corporation Publication,1969.23 “A Computer Program for the Analysis of Beam-Columns Under Static Axial and Lateral Loads,” Bogard, D. andH. Matlock, Offshore Technology Conference Paper 2953, 1977.


Chapter 10 Foundation Design

Contents:Section 1 — Vertical Resistance.........................................................................................10-3

Section 2 — Lateral Loads and Resistance.........................................................................10-6

Section 3 — Retaining Walls..............................................................................................10-7

Section 4 — Slope Stability..............................................................................................10-10

Chapter 10 — Foundation Design Section 1 — Vertical Resistance




Section 1 Vertical Resistance

Background

Variation in foundation material properties makes accurate design extremely complicated.There are many types and sub-types, ranging from soft soil to hard rock, that occur indifferent layers with properties that vary with moisture content and overburden. Mercifully,simple tests have been developed that allow structural engineers to estimate the verticalcapacity of most foundation elements with reasonable accuracy.

Early bridge designers relied heavily on spread footing foundations, although timber andconcrete piling were available options. Steel H piling became popular in the late 1930s. Afew caissons, pneumatic and open, were used for larger stream crossings. Drilled shafttechnology developed in the late 1940s. Drilled shaft and prestressed concrete pilefoundations now dominate for bridge construction in Texas.

Prior to 1940, bridge foundation matters were coordinated by the Plan Review Section of theBridge Division. Specialized expertise in general geotechnical engineering resided in theMaterials and Tests Division. A bridge foundation soils group was then formed to insurecontinuing and consistent handling of increasing foundation problems. One of the firstprojects of this group was development of a reliable soil test method for use withexploratory drilling rigs. With cooperation from the Materials and Tests Division and theEquipment and Procurement Division, the Texas cone penetrometer test was developed.During the following 15 years, shear strengths of various types of soil were correlated withthe number of blows of a 170 pound free-falling hammer required to drive the conepenetrometer a given distance into the founding material. Confidence in this test continuesto be justified by lack of foundation failures on Texas bridges.



Bridge Foundation Design Issues

Bridge foundations are a major factor in bridge design. Foundation and geotechnical issueare handled by the Texas Department of Transportation’s (TxDOT) geotechnical engineerswithin the Bridge Division. The geotechnical group that originated in plan review becamepart of the Construction Section of the Bridge Division in 1962. Around 1980 it became aseparate branch in the Bridge Design Section. Currently, geotechnical design andfoundations issues are the responsibility of the Geotechnical Branch of the Bridge Division -Technical Services Section.

The Geotechnical Branch performs the following:

♦ Coordinates preliminary exploration of foundation conditions

♦ Consults with bridge designers regarding type, size, and length of foundation elementsto be used

♦ Reviews plans prepared in the districts, or by consulting engineers, for properfoundation design

♦ Consults on construction problems associated with bridge foundations and retainingwalls.

Beginning in 1962, 11 research studies generating 37 reports on piling foundations weresponsored by TxDOT. The majority of these studies were performed by the TexasTransportation Institute at Texas A&M University.

Beginning in 1965, 10 research studies generating 24 reports on drilled shaft foundationswere sponsored by TxDOT. The majority were performed by the Center for TransportationResearch at the University of Texas.

Construction specifications dating from 1918 cover timber and concrete piling. Verticalload resistance was estimated from the last few hammer blows using the Engineering Newsformula. This approach is said to date back to 1888. Calculated resistance was required tobe equal to or greater than the pile load shown on the plans, which was calculated for serviceloads.

The same basic resistance formulas, modified for double acting power hammers, remain inthe current construction specifications. Additionally, dynamic analysis by a wave equationcomputer program is used, under some soil conditions, to determine safe load capacity. Thisprogram is also useful for computing stresses in the pile for various combinations ofhammer, cushion, and soil.

In 1965, the Texas Quick Load Test Method was incorporated into the specifications toestablish safe load capacity of piling and drilled shafts. One or more test loads are specifiedin rare instances where analytical design appears undependable and significant savings arepossible through more refined methods.

Test-driven piling are used more often to supplement the hammer formula or wave equationin verifying safety under the design load. For clayey soils, advantage may be taken of soilset up tendencies by redriving the test piling after seven days.



Drilled shaft usage was promoted vigorously by the industry and supported by TxDOTresearch. Point bearing design using bell footings was predominant. As confidence grewfrom research and test loads, skin friction was utilized to increase load resistance, and theuse of bell footings was virtually eliminated.

Verification of the reliability of the slurry displacement method, in the early 1970s, led toincreased usage of drilled shafts in previously questionable soil conditions. This methodalso eliminates the need for troublesome casing.

Current Status

For plans prepared in the Bridge Design Section, all foundation design is finalized by theGeotechnical Branch. Foundation designs prepared by district personnel and by consultingengineers are reviewed by the Geotechnical Branch.


If there is any doubt as to the type of foundation appropriate for a bridge, the GeotechnicalBranch should be consulted early in the design process. After the type of foundation hasbeen established, design should be completed and pile or drilled shaft loads calculated forservice loads. This information, with a print of the layouts and boring logs, should be givento the Geotechnical Branch to establish lengths and recommend special notes or testprocedures if needed.

Foundation design is accomplished using the information contained in theGeotechnical Manual.

Chapter 10 — Foundation Design Section 2 — Lateral Loads and Resistance


Section 2 Lateral Loads and Resistance

Background

Lateral loads imposed by soils and response of embedded elements to lateral loads are evenmore speculative than vertical load response. Classical methods of Rankine or Coulombhave been used for active pressure and passive resistance. TxDOT attempts at fieldmeasurement of lateral pressure and response confirm the high degree of variability. TheFederal Highway Administration (FHWA) Handbook on Design of Piles and Drilled Shaftsunder Lateral Load 1 is probably the most comprehensive treatment of the resistance ofsoils to lateral load.

Current Status

For most bridge designs, judgmental treatment of lateral load and resistance, based onexperience, is considered sufficient. For critical situations, computer programs are availablethat consider the effects of lateral soil resistance on the structural design.


For most conditions, lateral soil pressure may be assumed to be 40 pounds per square footper foot of height, including surcharge.

For questionable conditions, computer programs FRAME51 or COM624 may be used withlateral load-deflection values (p-y curves) as recommended by the geotechnical engineer.

For more complete information, see the Geotechnical Manual.

Chapter 10 — Foundation Design Section 3 — Retaining Walls


Section 3 Retaining Walls

Background

Although not strictly foundations, retaining walls are predominantly geotechnicallycontrolled.

For many years, retaining walls in Texas were mostly cantilever walls on spread footings.Buttress or counterfort walls were used occasionally. In extremely soft soils, footings wereplaced on piling of various types and configurations. Closely spaced drilled shafts havebeen used for crowded conditions since 1970. Later, prestressed ground anchors were addedto reduce the number and size of drilled shafts. In the late 1970s reinforced earth wallsbegan to be accepted and were often the contractor’s choice on alternate bids. Otherproprietary variations of mechanically stabilized earth (MSE) walls soon appeared. Sincethe mid-1980s proprietary retaining wall designs have proliferated.

Retaining walls were formerly designed in a Bridge Division or district structural group toresist active pressures calculated according to Rankine or Coulomb, or to 30 or 40 poundsper square foot per foot, as the designer saw fit. Beginning in 1971, TxDOT sponsored fiveresearch studies, generating 12 reports relative to retaining wall design. The latest report2

contains a design procedure for closely spaced drilled shaft walls. Geotechnical Branchengineers perform virtually all retaining wall designs for which the Bridge Design Section isresponsible, and review all others.

Various district standards for cantilever walls have been used. The Bridge Division issued acomprehensive set of cantilever retaining wall standards in 1975 and updated them in 1984,only to have them rendered virtually obsolete by the rise in popularity of MSE walls.Bidding information for MSE walls consists of one sheet of general design requirements,plan and elevation geometry for the walls and a construction specification. Detailed plansare prepared by the successful wall supplier.



Construction Issues

Most problems with retaining walls occur during construction. The most severe loadingcondition apparently occurs during backfilling. Observed malfunctions have been:

♦ Excessive wall deformation:• Excessive toe pressure due to soft foundation or poor backfill material and/or

procedures• Deformation of the wall itself due to heavy backfill equipment or deficient MSE

backfill

♦ Sliding of spread footing walls:• Insufficient vertical load such as with a short heel on cantilever walls• Loss of passive resistance by excavation in front of the toe

♦ Rotation and sliding of wall and footing:• General embankment stability failure

♦ Backfill failure:• Improper backfill materials for MSE• Inadequate compaction of backfill.

Maintenance Issues

Retaining wall malfunctions are very difficult to repair.

One notable time-related failure of a cantilever/spread footing wall occurred in San Antonio.Copper waterstops had been used between wall and footing. In a very soggy environment,the reinforcing steel became a sacrificial anode in a corrosion cell and graduallydeteriorated. Some 30 years after construction, several wall sections fell onto the highwayduring a rainstorm.

Current Status

The Geotechnical Branch of the Bridge Division - Technical Services Section is responsiblefor the design or review of all retaining walls on the state highway system.




For usual conditions, retaining walls may be designed for a pressure of 40 pounds per squarefoot per foot of wall height plus surcharge.

Closely spaced drilled shaft walls should be designed with the assistance of a lateral loadand resistance computer program such as COM624.

Current American Association of State Highway and Transportation Officials (AASHTO)Specifications contain a comprehensive section on retaining wall design, but this issubservient to the established practice of the Geotechnical Branch.

For specific guidance regarding retaining wall selection and design, refer to theGeotechnical Manual.

Chapter 10 — Foundation Design Section 4 — Slope Stability


Section 4 Slope Stability

Background

Slope failures apparently received very little attention prior to the interstate highway era.Either significant failures did not occur or failures were repaired without fanfare bymaintenance forces. Then, perhaps due to increasing construction intensity, greater cut andfill requirements, or public exposure, slope failures appeared to increase, reaching a peak inthe early 1980s with major failures in Houston, Corpus Christi, and Beaumont.

Slope failures are almost always associated with plastic clay soils and water. Many failuresoccur within cut slopes, but embankment slopes may also fail because of poor fill materialand penetration of surface water. Weak natural ground can allow the embankment failuresurface to extend below and considerably beyond the toe of the slope.

Migration of river banks is related to slope stability in that clay layers and ground water areusually involved. There have been several instances of bridge piers being severely displacedby movement of the surrounding soil toward the river.

Beginning in 1971, TxDOT sponsored four research studies generating nine reports relativeto slope stability. One of these3 contains a survey of slope failures and repair methods inTexas.

Bridge engineers are not often involved in slope stability design until instability threatens abridge or retaining wall. When such failures occur, the cost of repair is great. Bridgeengineers are often asked to participate in the design of structures to repair local slopefailures not associated with a retaining wall. Mitigation of the effects of river bankmigration taxes the ingenuity of bridge and geotechnical engineers alike.

Current Status

Highway design engineers are generally responsible for initial evaluation of slope stability.The Geotechnical Branch is responsible for slope stability when bridges or retaining wallsare likely to be involved, or when otherwise asked to evaluate potential problem sites.


Bridge engineers should not hesitate to bring questionable slope conditions to the attentionof the Geotechnical Branch. Experience, computer programs, and other analytical methodsare available for proper consideration of the problem.

For specific guidance regarding slope stability issues, refer to the Geotechnical Manual. 1 “Handbook on Design of Piles and Drilled Shafts Under Lateral Load, Turner Fairbank Highway Research Center,Report No. FHWA-IP-84-11, 1984.

Chapter 10 — Foundation Design Section 4 — Slope Stability


2 “Study of Design Method for Vertical Drilled Shaft Retaining Walls,” Wang, S.T. and L.C. Reese, CTR, FinalReport 415-2F, 1985.3 “A Survey of Earth Slope Failures and Remedial Measures in Texas,” Abrams, T.G. and S.G. Wright, CHFR,Report 161-1, 1972.

Bridge Design Manual A-1 TxDOT 12/2001

Appendix ASupplemental Design Criteria for Prestressed Concrete U

Beams

U-Beam Spacing

Use the following guidelines for the spacing of U beams.

U-beam Spacings vs. Span LengthsSpan Length (ft.) Maximum Beam Spacing (ft.)

U40 Beam U54 Beam75 16.08 16.6780 16.08 16.6785 15.75 16.6790 14.00 16.6795 12.50 16.67100 11.00 16.67105 9.50 16.67110 7.50 15.25115 NA 13.75120 NA 12.25125 NA 11.00130 NA 9.75

Guidelines in the preceding table assume the following design criteria:

♦ Interior beam design only

♦ 8" slab thickness (maximum clear span = 8'-8")

♦ Maximum composite width to either side of top flange = 48"

♦ Live load distribution factor = S/11 per truck/lane, with a minimum value of 0.9

♦ Maximum debonding length = 0.2L or 15', whichever is less

♦ 75% maximum debonding per row, 75% maximum debonding per section

♦ 0.110 klf composite dead load (1/3rd of T501 rail ~ 0.330 klf)

♦ 50% relative humidity (used only for producing table)

♦ 1/2" 270 ksi low-relaxation strand

♦ f'ci max = 6500 psi, f'c max = 8500 psi

♦ No overlay

♦ Span lengths shown are CL to CL Bent with 9 ½" distance to CL Bearing

Appendix A Supplemental Design Criteria for Prestressed Concrete U Beams


Bearing Pad Taper Calculations for U Beams

The bearing seat for a U beam is level perpendicular to the centerline of bearing but slopesalong the centerline of bearing between the left and right bearing seat elevations. Thebearing pad is oriented along the centerline of bearing. This configuration for the U-beambearing allows the pad to taper in only one direction (perpendicular to the centerline ofbearing). The amount of bearing pad taper depends on three factors - the grade of the Ubeam, the slope of the bearing seat, and the beam angle.

All U-beam jobs should have a Bearing Pad Taper Report sheet that contains the various padtapers for use by the bearing pad fabricator. This report summarizes the bearing pad taperperpendicular to the centerline of bearing for each beam bearing location. In RDS, the reportis titled “Bearing Pad Taper -- Fabricator’s Report”. The calculations that follow derive theformula for calculating the bearing pad taper perpendicular to centerline of bearing.

For purposes of developing the formulas for calculating bearing pad taper, the followingsign convention will be used looking in the direction of increasing station numbers:

♦ positive bearing seat slope is up and to the right

♦ negative bearing seat slope is down and to the right

♦ positive beam grade is up

♦ negative beam grade is down

♦ positive bearing pad taper is up

♦ negative bearing pad taper is down



CASE I. Beam Angle is θ < 90° and Right Forward.

Figure A-1. Plan View of Bearing Seat with Right Forward Beam Angle. Online users canclick here to view this illustration in PDF format.

Perpendicular to the centerline of bearing, the pad taper is only function of the bottomsurface of the beam. The bearing seat is level in this direction, so the component of pad taperdue to the top surface of the bearing seat is zero. Using ELEV1, ELEV2, and ELEV3 atpoints 1, 2, and 3, respectively, at the bottom of U beam, the equation for pad taper is:

TAPER = (ELEV2 – ELEV3)/W

Where:

ELEV2 = ELEV1 + BEAM GRADE × (W/sin θ)

ELEV3 = ELEV1 + SLOPE × (W/tan θ)

Substituting for ELEV2 and ELEV3 , the equation for pad taper becomes:

TAPER = (BEAM GRADE – SLOPE × cos θ)/sin θ

CASE II. Beam Angle is θ < 90° and Left Forward.



Figure A-2. Plan View of Bearing Seat with Left Forward Beam Angle. Online users canclick here to view this illustration in PDF format.

Using ELEV1, ELEV2, and ELEV3 at points 1, 2, and 3, respectively, at the bottom of the Ubeam, the equation for pad taper is:

TAPER = (ELEV2 – ELEV3)/W

Where:

ELEV2 = ELEV1 + BEAM GRADE × (W/sin θ)

ELEV3 = ELEV1 – SLOPE × (W/tan θ)

Substituting for ELEV2 and ELEV3, the equation for pad taper becomes:

TAPER = (BEAM GRADE + SLOPE × cos θ)/sin θ

CASE III. Beam Angle is θ = 90°

Since the centerline of the U beam is perpendicular to centerline of bearing, the componentof bearing pad taper due to the bearing seat is zero, and the bearing pad taper is simply:

TAPER = BEAM GRADE



Summary

Figure A-3. Plan View of Bearing Seat. Online users can click here to view this illustrationin PDF format.

Defining β = beam angle as measured counterclockwise from centerline of bearing, theequation for the calculating bearing pad taper for any bearing location is as follows:

TAPER = (BEAM GRADE – SLOPE × cos β)/sin β



Figure A-4. Example Shear Key Detail. Online users can click here to view this illustrationin PDF format.

Haunch Calculations for U Beams

U beams are not placed vertical like I beams but at a cross-slope. For spans with constantcross-slope and constant overall width, U beams will be at the same cross-slope as theroadway surface. For spans with more complicated geometry, such as varying cross-slopeand/or varying overall width, U beams will be at some cross-slope other than the cross-slopeof the roadway surface. Each U beam in a span is balanced in cross-slope from the backbearing to the forward bearing of the beam so that no torsion is introduced into the beam.Thus, the haunch at centerline of bearing for the left edge of the beam may be different thanthe haunch at right edge of the beam. Skewed beam end conditions can also contribute to adifferent haunch at centerline bearing for each edge of the beam (this difference can existeven with a constant cross-slope, i.e., it is due to the geometry of the roadway surface andnot necessarily the balancing of the U beam).

In terms of calculating the required haunch at centerline of bearing for a U beam, the haunchfor each edge of top flange of the U beam must be calculated. Once the minimum haunchvalue is established, the maximum haunch at centerline of bearing on the opposite top edgeof the U beam can be calculated as well as the deduct value for computing bearing seatelevations. The following is a suggested method of calculating the required haunch and thecorresponding deduct values for U beams:



Step 1. Execute a preliminary RDS run using the beam framing options 20, 21, or 22 inorder to calculate the vertical curve component of the haunch. On the BRNG card, input thesection depth as zero and the pedestal width equal to the top flange width of the U beam(7.42' for U40 beams and 8.00' for U54 beams). Be sure to add the letter “P” in column 80 ofthe BRNG card. This will instruct RDS to keep the top flange width dimensionperpendicular to the centerline of U beam. Thus, for skewed beam end conditions, RDS willtake into account the skew at that end of the U beam and give the corresponding verticalordinates at the centerline of bearing for the left and right edges of the top flange (See FigureA-5). Also, include a VCLR card for each span with the bridge alignment as the specifiedalignment.

Step 2. Examine the RDS output. Three lines of vertical ordinates will be generated forevery U beam, i.e., the vertical ordinates along the left top edge, centerline, and right topedge of U beam. The first and last columns of each vertical ordinate table are the ordinatesat centerline of back bearing and forward bearing, respectively. One or both vertical ordinatevalues at the left and right top edge of the U beam at centerline of bearing will be zero. Avertical ordinate of zero indicates that the top edge of the U beam is matched with theelevation of the top of slab at that point. Thus, the vertical placement is controlled by these“corners” of the U beam that have vertical ordinates of zero.

The corner opposite to the controlling corner at the centerline of bearing will either have azero or negative vertical ordinate. Its value depends on the bridge geometry and/or balancingof the U beam. A zero value for the vertical ordinate at the dependent corner means that atthat point the top of the U beam is also matched with the elevation of the top of slab.However, a negative vertical ordinate value at the dependent corner means that at that pointthe top edge of U beam is below the elevation at top of slab. This negative vertical ordinateis the difference in haunch at centerline of bearing from the left top edge of beam to the righttop edge of beam due to bridge geometry and/or balancing of the U beam. (See Figure A-5).

Figure A-5. Example Plan View of the Vertical Ordinates for U Beam. Online users canclick here to view this illustration in PDF format.



Figure A-6 illustrates the vertical ordinates produced by RDS for the left and right top edgesof the U beam when framing a span with a crest vertical curve and a varying cross-slopealong the span. It is shown only to help visualize a possible scenario of vertical ordinatesproduced by RDS.

Figure A-6. Example Vertical Profile at Edges of U Beam. Online users can click here toview this illustration in PDF format.



Figure A-7. Example of U Beam Geometry as per RDS. Online users can click here to viewthis illustration in PDF format.

When inputting the top flange width of the U beam, bf, on the BRNG card, the VCLRcommand calculates the vertical ordinates at an offset distance of bf/2 from the RDS beamline (See Figure A-7.). The standard convention for defining the RDS beam line is a verticalline at a point coinciding with the centerline of the bottom of the bearing pad. Thus, for Ubeams at a cross-slope, the beam rotates about this point. This rotation of the U beam shiftsthe top flange of the beam transversely with respect to the RDS beam line. RDS makes noadjustment for the rotation-induced transverse movement and, therefore, yields verticalordinate calculations that are not exactly at the outside edge of the top flange. However, thiserror should be negligible as the offset error will only be approximately 1" for a U54 on a2% cross-slope.

Step 3. Calculate the required minimum haunch at centerline bearing that will work for allU beams in a span. Start by calculating the required minimum haunch at centerline ofbearing for both the left and right top edges of each U beam in that span. For each side of thebeam, work from the controlling corner and use the entire maximum vertical ordinate on thatedge in your haunch calculation. Do not be concerned with the vertical ordinate value at thedependent corner for each side because that value affects only the maximum haunch (seeStep 4), not the minimum haunch. Also, we typically use 75% of the predicted camber byPrestress 14 for U beams because in the field we have not been consistently getting ourpredicted cambers. Keeping the sign convention used by RDS [a positive vertical ordinatemeans the top of beam is above the top of slab at that point, while a negative ordinate meansthe top of beam is below the top of slab at that point], the required minimum haunch valuesat centerline of bearing for each U beam will be:

Left Top Edge: Min. Haunchreq’d = 0.75C - 0.8∆DL + VOmax L + Min. HaunchCL Span

Right Top Edge: Min. Haunchreq’d = 0.75C - 0.8∆DL + VOmax R + Min. HaunchCL Span

Where:

VOmax L = maximum vertical ordinate, left top edge (usually at mid-span)

VOmax R = maximum vertical ordinate, right top edge (usually at mid-span)

C = camber of U beam

�DL = dead load deflection of U beam due to slab only



Min. HaunchCL Span = minimum haunch specified at centerline span (usually ½")

Using the largest required haunch value for that span:

Min. Haunchused = Min. Haunchreq’d (round up to the nearest ¼")

This haunch value will be the haunch at all controlling corners for each U beam in that span.

Step 4. Calculate the corresponding maximum haunches at centerline of bearing. Themaximum haunches at centerline of bearing occur at the dependent corners of each U beam.These maximum haunches may vary between U beams in a span but typically will not varyfor the same U beam. The maximum haunches at centerline of bearing for each U beam in aspan are:

Left Top Edge: Max. Haunchcal’d = Min. Haunchused - VOLt Depdt Corner

Right Top Edge: Max. Haunchcal’d = Min. Haunchused - VORt Depdt Corner

Where:

VOLt Depdt Corner = vertical ordinate value, left top edge, dependent corner

VORt Depdt Corner = vertical ordinate value, right top edge, dependent corner

Step 5. Calculate the slab dimensions at centerline of bearing, Xmin and Xmax, and thetheoretical slab dimensions at mid-span, ZL and ZR, for each U beam in the span (See FigureA-8). The equations are:

Xmin = Min. Haunchused + slab thickness

Xmax = Max. Haunchcal’d + slab thickness

ZL = Min. Haunchused - Min. Haunchreq’d @ left edge + slab thickness + Min. HaunchCL Span

ZR = Min. Haunchused - Min. Haunchreq’d @ right edge + slab thickness + Min. HaunchCL Span

Again, Xmin is the section depth at all controlling corners of the beam while Xmax is thesection depth at all dependent corners of the beam (any difference in Xmax for eachdependent corner of an individual beam should be negligible).



Figure A-8. Section Depth Information on Production Drawings. Online users can clickhere to view this illustration in PDF format.

Figure A-8: above shows a typical plan view and table that can be used on productiondrawings to describe the depth and location of the X and Z values. In order to use a singleand generic detail, the “min” and “max” designation is changed to an open convention usingthe letters “A” and “B”. As a result, XA and XB can be either the Xmin or Xmax values.

Step 6. Calculate the required deduct at the specific bearing location to use in computing thefinal bearing seat elevations. The first table in this appendix shows the pedestal widths forthe U40 and U54 beams with the standard and dapped end conditions. The pedestal widthslisted depend on the beam angle and are adequate for up to two 9" x 19" bearing pads. Also,because you can only input one pedestal width per BRNG card, the pedestal width usedmust be for the U beam with the smallest beam angle at that bearing location. Be sure toomit the letter “P” in column 80 of the BRNG card so that RDS applies the pedestal widthalong the centerline of bearing.

Standard Bearing Seat Dimension “D”Beam Angle Standard End Dapped End

75�� θ � 90� 4'-6" 5'-0"60�� θ < 75� 5'-0" 5'-6"45�� θ < 60� 5'-6" 6'-0"



Figure A-9. Location of Deduct for Final Bearing Seat Elevations. Online users can clickhere to view this illustration in PDF format.

The required deduct for calculating bearing seat elevations needs to be the deduct at the edgeof the bearing seat (see Figure A-9.) This deduct can be obtained by interpolating betweenthe values Xmin and Xmax for each U beam using the beam angle, the top flange width of thebeam, and the chosen bearing seat width. The largest calculated deduct at that bearinglocation should be used to compute the final bearing seat elevations for all the U beams atthat bearing location. The difference between the calculated deducts at a given bearinglocation should be negligible, but you may want to check your worst case span initially tosee if the difference is large enough to take into account The deduct for the calculation offinal bearing seat elevations is:

Deduct = (Xmax - Xmin)/(bf /sin θ) x (bf - D)/2 + Xmin + Beam Depth + Brng. Pad Thickness

Where:

bf = top flange width

θ = beam angle

D = chosen pedestal width

Note: The beam angle can probably be ignored because of negligible difference in the finaldeduct amount. In addition, this formula does not include the vertical adjustment of thebeam depth and bearing pad thickness due to the cross-slope of the beam, which is alsonegligible.

Summary

The required haunch at centerline of bearing for the left and right top edges of the U beamshould always be calculated working from the controlling corner for that side. This is donebecause the vertical ordinate for the dependent corner is “built-in” to the geometry for thebeam and bridge. We cannot use that value in determining our haunch because the verticalordinate at the dependent corner is always present, i.e., we cannot adjust the beam verticallyto reduce that dimension. Basically, the controlling corners will have the minimum haunchat centerline of bearing while the dependent corners will have the maximum haunch atcenterline of bearing, the difference being the vertical ordinate value at the dependent



corner. Incidentally, because the U beam is at an average cross-slope, the haunches atcenterline of bearing for the back end of the beam for the left and right edges will typicallybe reversed at the forward end of the beam.

At mid-span, the theoretical haunch value for the left and right top edges of each U beamwill be the same value if the roadway surface cross-slope is constant or transitions at aconstant rate over the entire length of the span and the beam spacing remains constant in thespan. For any other case, the theoretical haunch at mid-span may be different for the leftand right top edges of each U beam.

Bridge Design Manual B-1 TxDOT 12/2001

Appendix BDesign Formulas for Inverted Tee Bents

Introduction

In addition to conventional design requirements for flexure, shear, and side beamreinforcement, the following design recommendations for ledge depth, ledge reinforcement,and web reinforcement are given. This information is adapted from material published inDesign of Reinforced and Prestressed Concrete Inverted T-Beams for Bridge Structures.1

Appendix B Design Formulas for Inverted Tee Bents

Bridge Design Manual B-2 12/2001

Figure B-1. General Information. Online users can click here to view this illustration inPDF format.



Ledge Depth

Figure B-2. Punching Shear. Online users can click here to view this illustration in PDFformat.

Punching Shear: .85=φ

Note: These equations are for rectangular pads with no skew. However, these equationscan generally serve as a basis for determining punching shear capacity for other cases.

Interior Beam: pvpvn ddZWBcfP )222(4 +++′= φφ

Exterior Beam: pvpvn dCdZWBcfP )5(.4 ++++′= φφ , IfpvdZWBC +++≤ 5.

Shear Friction: .85=φ

Note: Use a maximum f ′c of 4000 psi. This limit is imposed because when computing Avfto determine required ledge reinforcing the formulas given would becomeunconservative for higher concrete strengths. Since punching shear is usually morecritical than shear friction in determining ledge depth this limitation seldom affects finaldesign.

) (2. sfn dwidthdistrcfP ′= φφ

Distribution width equals the lesser of:

Interior Beam: SaB or 4+

Exterior Beam: SCC 5.or 2 +

Note: If )4(5. aBC +> use interior beam criteria



Ledge Reinforcing

Figure B-3. Ledge Reinforcing. Online users can click here to view this illustration in PDFformat.

Shear Friction Requirement: 85.=φ

y

uvf

fPA

4.1×=

φ


Interior Beam: SaB or )4( +

Exterior Beam: )4(or ),5.(,2 aBSCC ++

Flexure Requirement: 9.=φ

sfy

usf

dfaPA8.0××

×=φ

, but not less than widthdistfdcf

y

sf .04. ××′






Tension Requirement: 9.=φ

uuy

un PN

fNA 2. where, =×

=φ




Top Layer:

widthdistrA

widthdistrA

sA nvf

. .667. +≥ , or

widthdistrAA nsf

.+

Second Layer:

widthdistrA

sA vf

.333.≥



Web Reinforcing

Hanger Reinforcement: 85.=φ

Design for the largest value of sAv from the following equations.



)2(

42

pvy

pvfu

v

dBf

dbcfP

sA

+

××′−= φ

SfP

sA

y

uv

φ2= for interior beams or

CfP

sA

y

uv

22×

=φ

for exterior beams, if 2SC <

)3(3

aBfP

sA

y

sv

+= for interior beams or

)2(3CfP

sA

y

sv = for exterior beams, if 2

3 aBC +<

Where:

spacing stirrup legs stirrup both/all of Area

cap theof side oneon reaction beam single load servicelargest The cap theof side oneon reaction beam single factoredlargest The

====

sAPP

v

s

u

Note: It is recommended that the distance from centerline of exterior beam at centerline ofbearing to the end of cap (C) should not be less than 24 inches. This allows for areasonable size and spacing of hanger reinforcement and limits cracking between theflange and stem. Also, additional hangers with anchorage hooks at their ends should bearbitrarily placed along the end face of inverted-T caps. These end hangers should besized to match the stirrups and spaced at 6 inches max.

Shear: 85.=φ

)2( wyv

wwn dfsAdbcfV +×′= φφ

Note: Take Vu at face of column, not depth of cap from column face. Stirrup requirementsfor shear are normally not added to the hanger requirements. (See Appendix Areference 56)

Torsion: 85.=φ

Figure B-4. Torsion Elements. Online users can click here to view this illustration in PDFformat.

( )�

��

�+

′= 11

2

23

4yxf

sAyxcf

T tyv

n αφφ , but �

��

�

�′≤

318

2yxcfφ



Where 5.133.66.1

1 ≤+=xy

tα

0.122

≤�

��

�+��

��

�

n

u

n

u

TT

VV

φφ

If >1.0 add additional stirrups sA v′

and additional longitudinal steel ( )11 yxsA

Av

+′

=

Note: A should be distributed to the four corners of the web, the four corners of the beamledges, and added to the flexural reinforcing.

1 “Design of Reinforced and Prestressed Concrete Inverted T-Beams for Bridge Structures,” Furlong and Mirza, PCIJournal, July – August 1985.

Bridge Design Manual C-1 TxDOT 12/2001

Appendix CLoad vs. Moment Interaction Diagrams

Introduction

The following Load vs. Moment interaction diagrams were developed for 24- through 72-inch diameter, round columns with common reinforcing patterns used on TxDOT projects.Diagrams for 3000 psi and 3600 psi concrete represent commonly used drilled shaft andcolumn concrete strengths. Note that the reinforcing is grade 40. This is because grade 40 iscommonly used in the design of interior bent columns even though grade 60 will be requiredby the specifications. Column design is normally not a strength issue. The AASHTOminimum reinforcement ratio of one percent is normally the controlling design feature.Using grade 40 allows for considerably shorter lap and embedment requirements, reducingcongestion and improving constructibility.

Appendix C Load vs. Moment Interaction Diagrams


Figure C-1. Diagrams for 24-Inch Diameter Columns. Online users can click here to viewthis illustration in PDF format.



Figure C-9. Diagrams for 72-Inch Diameter Columns. Online users can click here to viewthis illustration in PDF format

bridge design manual-texas department of transportation

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