1

GEOTECHNICAL SUMMARIESGeneral

We will be using the term foundation to describe the structural elements that connect a structure to the ground. These elements are made of concrete, steel, wood, or perhaps other materials. We will divide foundations into two broad categories :

1. Shallow foundations Shallow foundations transmit the structural loads to the near-surface soils.

2. Deep foundationsDeep foundations transmit somes or all of the loads to deeper soils.Classification of foundations are ilustrated in this figure belows :

Figure 3. Types of foundationCommon types of piles are as follows :

Batter pile

A pile driven in at an angle inclined to the vertical to provide high resistance to lateral loads.

End-bearing pile

A pile whose support capacity is derived principally from the resistance of the foundation material on whinch the pile tip rests. End-bearing piles are often used when a soft upper layer is underlain by a dense or hard strata. If the upper soft layer should settle, the pile could be subjected to down-drag forces, and the pile must be designed to resist these soil-induced forces.

Friction pile

A pile whose support capacity is derived principally from the resitance of the soil friction and/or adhesion mobilized along the side of the pile. Friction piles are often used in soft clays where the end-bearing resistance is small because of punching shear at the pile tip.

Combined end-bearing and friction pile

A pile that derives its support capacity from combined end-bearing resistance developed at the pile tip and friction and/or adhesion resistance ont pile perimeter.

Once the design loads have been defined, we need to develop foundation designs that satisfy several performance requirements. The first category is strength requirements, which are intended to avoid catastrophic failures. There are two types : Geotechnical strength requirements

Geotechnical strength requirements are those that address the ability of the soil or rock to accept the loads imparted by the foundation without failing. The strength of soil is governed by its capacity to sustain shear stresses, so we satisfy geotechnical strength requirements by comparing shear stresses with shear strengths and designing accordingly.

Geotechnical strength analysis are almost always performed using allowable stress design (ASD) methods.

Structural strength requirements

Structural strength requirements address the foundations structural integrity and its ability to safely carry the applied loads. Corrosion of Steel

Under certain conditions, steel can be the object of extensive corrosion. This can be easily monitored when the steel is above ground, and routine maintenance, such as painting, will usually keep corrosion under control. However, it is impossible to inspect underground steel visually, so it is appropriate to be concerned about its potential for corrosion and long-term integrity.

For corrosion assessment, steel foundations can be divided into two categories: those in marine environments and those in land environments. Both are shown in Figure below :

Figure 4. (a) Marine environments include piers, docks, drilling platforms, and other similar structures where a portion of the foundation is exposed to open water. (b) Land environments include buildings and other structures that are built directly on the ground and the entire foundation is buried.Safety Factor of Foundation Bearing CapacityMost building codes do not specify design factors of safety. Therefore, engineers must use their own discretion and proffesional judgment when selecting F. Items to consider when selecting a design factor of safety include the following :

Soil type

Shear stregth in clays is less reliable than that in sands, and more failures have occured in clays than in sands. Therefore, use higher factors of safety in clays.

Site characterization data

Projects with minimal subsurface exploration and laboratory or in-situ tests have more uncertainty in the design soil parameters, and thus require higher factor of safety. However, when extensive site characterization data is available, thier is less uncertainty so lower factors of safety may be used. Importance of the structure and the consequences of a failureImportant projects, suc as hospitals, where foundation failure would be more catstrophic may use higher factors of safety than less important projects, suc as agricultural storage buildings, where cost of construction is more important. Likewise, permanent sturctures justify higher factors of safety than temporary sturctures, such as construction falsework. Structures with large height-to-width ratio, such as chimneys or towers, could experience more catastrophic failure, and thus should be designed using higher factors of safety.

The likelihood of the design load ever actually occuringSome structures, such as grain silos, are much more likely to actually experience their design loads, and thus might be designed using a higher factor of safety. Conversly, office buildings are much likely to experience the design load, and might use a slightly lower factor of safety.The true factor of safety is probably much greater than the design factor of safety, because of the following :

The shear strength data are normally interpreted conservatively, so the design values of c and ( implicitly contain another factor of safety.

The service loads are probably less than the design loads.

Settlement, not bearing capacity, often controls the final design, so the footing will likely be larger than that required to satisty bearing capacity criteria.

Spread footings are commonly built somewhat larger than the plan dimensions.

Safety Factor for Bearing Capacity of Single PileRefers to Engineering Manual (EM 1110-2-2906) :Table 1. Safety Factors for Bearing Capacity of Pile FoundationMethod of Determining CapacityLoad ConditionMinimum Factor of Safety

CompressionTensile

Theoretical of empirical prediction to be verified by pile load testUsual

Unusual

Extreme2,0

1,5

1,152,0

1,5

1,15

Theoretical or empirical prediction to be verified by pile driving analyzerUsual

Unusual

Extreme2,5

1,9

1,43,0

2,25

1,7

Theoretical or empirical prediction not verified by load testUsual

Unusual

Extreme3,0

2,25

1,73,0

2,25

1,7

Loading conditions : Usual. These conditions include normal operating and frequent flood conditions. Basic allowable stresses and safety factors should be used for this type of loading condition.

Unusual. Higher allowable stresses and lower safety factors may be used for unusual loading conditions such as maintenance, infrequent floods, barge impact, construction, or hurricanes. For these conditions allowable stresses may be increased up to 33 percent. Lower safety factors for pile capacity may be used, as described in Table above.

Extreme. High allowable stresses and low safety factors are used for extreme loading conditions such as accidental or natural disasters that have a very remote probability of occurrence and that involve emergency maintenance conditions after such disasters. For these conditions allowable stresses may be increased up to 75 percent. Low safety factors for pile capacity may be used as described in Table above. An iterative (nonlinear) analysis of the pile group should be performed to determine that a state of ductile, stable equilibrium is attainable even if individual piles will be loaded to their peak, or beyond to their residual capacities. Special provisions (such as field instrumentation, frequent or continuous field monitoring or performance, engineering studies and analyses, constraints on operational or rehabilitation activities, etc.) are required to ensure that the structure will not catastrophically fail during or after extreme loading condition.

Safety Factor of Pullout Capacity

Table below gives typical design factors of safety based on these assessments. Consider them to be guides, not absolute dictates, so do not hesitate to modify them as necessary.Table 2. Safety Factor of Pullout Capacity

Classification of StructureAcceptable Probability of FailureDesign Factor of Safety, F

Good ControlNormal ControlPoor ControlVery Poor Control

Monumental10-52,33,03,54,0

Permanent10-42,02,52,83,4

Temporary10-31,42,02,32,8

Safety factor of pullout capacity will be equal of 3,00Safety Factor of Lateral Resistance

The definition of failure load should therefore be related to the acceptable or tolerate lateral deformation of the structure. Where no such criteria are available, 0.25 in. (6.25 mm) is considered as the criterion on which failure load is established. It should be realized that actual instability at which the load could not be held when the pile head had deformed about 1 in. (25 mm). (refers to Pile Foundations In Engineering Practice, Shamsher Prakash and Hari D. Sharma)

For lateral resistance design of pile foundation, we use some boundary :

Table 3. Maximum Design for Lateral Deflection of Pile Foundation LoadMaximum Lateral Defflection

Service Load without earthquake load0,625 cm (0,25)

Service Load with earthquake load1,25 cm (0,50)

Ultimate Load2,50 cm (1,00)

Allowable Settlement

The amount of settlement that a foundation can tolerate is called the allowable settlement, the magnitude of this settlement depends upon it mode.

A structure that has undergone uniform settlement is one where all points within the structure have moved vertically the same amount (Figure a). This type of settlement does not result in structural damage if it is constant across whole structure. However, there will be problems with appurtenances such as with pipes, entrance-ways, etc. Another possibility is settlement that varies linearly across the structure as shown in (Figure b). This causes the structure to tilt. Finally, (Figure c) shows a structure with irregular settlements. This mode distorts the structure and typically is the greatest source of problems.

Figure 5. Modes of settlement; (a) Uniform; (b) Tilting with no distortion; (c) Distortion

Figure 6. Types of foundation settlementTable below presents typical design values for the allowable settlement, (a. The design meets total settlement requirements if the following condition is met :

Where :

( = total settlement of foundation

(a = allowable settlementTable 4. Allowable Settlement (Sowers, 1962)Type of MovementLimiting FactorLimiting Factor

Total SettlementDrainage

Access

Probability of non-uniform settlement :

- Masonry walled structure

- Framed structures

- Smokestacks, silos, mats15-30 cm (6-12 in.)

30-60 cm (12-24 in.)

2,5-5 cm (1-2 in.)

5-10 cm (2-4 in.)

8-30 cm (3-12 in.)

TiltingStability against overturning

Tilting of smokestacks, towers

Rolling of trucks, etc.

Stacking of goods

Machine operation cotton loom

Machine operation turbogenerator

Crane rails

Drainage of floorsDepends on H and W

0,004L

0,01L

0,01L

0,003L

0,0002L

0,003L

0,01-0,02L

Differential MovementHigh continuous brick walls

One-story brick mill building, wall cracking

Plaster cracking (gypsum)

Reinforced-concrete-building frame

Reinforced-concrete-building curtain walls

Steel frame, continuous

Simple steel frame0,0005-0,001L

0,001-0,002L

0,001L

0,0025-0,004L

0,003L

0,002L

0,005L

Note :

L = distance between adjacent columns that settle diffrent amounts

H = height of structure

W = width of structure

- Higher values are for regular settlements and more tolerant structures.

- Lower values for irregular settlement and critical structures.

Table below also presents typical design values for the allowable total settlement, (a. These values already include a factor of safety, and thus may be compared directly to the predicted settlement.Table 5. Typical Allowable Total Settlements for Foundation Design Type of StructureTypical Allowable Total Settlement, (a

(in)(mm)

Office buildings0,5-2,0 (1,0 is the most common value)12-50 (25 is the most common value)

Heavy industrial buildings1,0-3,025-75

Bridges2,050

Diffrential Settlement

Engineer normally design the foundations for a structure such that all of them have the same computed total settlement. Thus, in theory, the structure will settle uniformly. Unfortunately, the actual performance of the foundations will usually not be exactly as predicted, with some of them settling more than expected and other less. This discrepancy between predicted behavior and actual behavior has many causes, including the following :

The soil profile may not be uniform across the site. This is nearly always true, no matter how uniform it might appear to be.

The ratio between the actual load and the design load may be different for each column. Thus, the column with the lower ratio will settle less than that with the higher ratio.

The ratio of dead load to live load may be different for each column. Settlement computations are usually based on dead-plus-live load, and the foundations are sized accordingly. However, in many structures much of the live load will rarely, if ever, occur, so foundations that have a large ratio of design live load to design dead load will probably settle less than those carrying predominantly dead loads.

The as-built foundation dimensions may differ from the plan dimensions. This will cause the actual settlements to be correspondingly different.

The differential settlement, (Da, is the difference in total settlement between two foundations or between two points on a single foundation. Differential settlements are generally more troublesome than total settlements because they distort the structure. This causes cracking in walls and other members, jamming in doors and windows, poor aesthetics, and other problems. If allowed to progress to an extreme, differential settlements could threaten the integrity of the structure.Therefore, we define a maximum allowable differential settlement, (Da, and design the foundation so that :

Table below presents a synthesis of these studies, expressed in terms of the allowable angular distortion, (Da. These values already include a factor safety of at least 1.5, which is why they are called allowable. We use them to compute (Da as follows :

Where:

(Da= allowable differential settlement

a= allowable angular distortion (from table below)L= distance between adjacent columns that settle different amountsTable 6. Allowable Angular Distortion, a (Wahls, 1994; AASTHO, 1996; and Other Sources)Type of Structurea

Steel tanks1/25

Bridges with simply-supported spans1/125

Bridges with continous spans1/250

Buildings that are very tolerant of differential settlements, such as industrial buildings with corrugated steel siding and no sensitive interior finishes1/250

Typical commercial and residential buildings1/500

Overhead traveling crane rails1/500

Buildings that are especially intolerant of differential settlement, such as those with sensitive wall or floor finishes1/1000

Machinery1/1500

Buildings with unreinforced masonry load-bearing wallsLength/height 3

Length/height ( 51/2500

1/1250

Site Investigation

The site investigation phase of the exploration program consists of planning, making test boreholes, and collecting soil samples at desired intervals for subsequent observation and laboratory tests. The approximate required minimum depth of the borings should be predetermined. The depth can be changed during the drilling operation, depending on the subsoil encountered. To determine the approximate minimum depth of boring, engineers may use the rules established by the American Society of Civil Engineers (1972):

Determine the net increase in the effective stress, ((, under a foundation with depth as shown in figure. Estimate the variation of the vertical effective stress, (0, with depth.

Determine the depth, D = D1, at which the effective stress increase (( is equal to 0,1q (q = estimated net stress on the foundation).

Determine the depth, D = D2, at which ((/(0 = 0,05.

Choose the smaller of the two depths, D1 and D2, just determined as the approximate minimum depth of boring required, unless bedrock is encountered.

Figure 7. Determination of the minimum depth of boring

If the preceding rules are used, the depths of boring for a building with a width of 30 m (100ft) will be approximately the following, according to Sowers and Sowers (1970):

Table 7. Boring depth for buildingNo. of storiesBoring depth

13,5 m (11 ft)

26 m (20 ft)

310 m (33 ft)

416 m (53 ft)

524 m (79 ft)

To determine the boring depth for hospitals and office buildings, Sowers and Sowers also use the rule

(for light steel or narrow concrete buildings)

And

(for heavy steel or wide concrete buildings)Where:

Db= depth of boring, in meters

S= number of stories

When deep excavations are anticipated, the depth of boring should be at least 1,5 times the depth of excavation.Sometimes, subsoil conditions require that the foundation load be transmitted to bedrock. The minimum depth of core boring into the bedrock is about 3 m. If the bedrock is irregular or weathered, the core borings may have to be deeper.

There are no hard-and-fast rules for borehole spacing. Table below gives some general guidelines. Spacing can be increased or decreased, depending on the condition of the subsoil. If various soil strata are more or less uniform and predictable, fewer boreholes are needed than in non-homogeneous soil strata.

Table 8. Approximate spacing of boreholesType of projectSpacing

Multistory building10-30 m (30-100 ft)

One-story building20-60 m (60-200 ft)

Highways250-500 m (800-1600 ft)

Residential subdivision250-500 m (800-1600 ft)

Dams and dikes40-80 m (130-260 ft)

The engineer should also take into account the ultimate cost of the structure when making decisions regarding the extent of field exploration. The exploration cost generally should be 0,10,5% of the cost of the structure. Soil borings can be made by several methods, including auger boring, wash boring, percussion drilling, and rotary drilling.

Auger boring is the simplest method of making exploratory boreholes. There are two types of hand auger: the posthole auger and the helical auger. Hand augers cannot be used for advancing holes to depths exceeding 3-5 m. However, the can be used for soil exploration work on some highways and small structures. Portable power-driven helical augers (76 mm to 305 mm in diameter) are available for making deeper boreholes. The soil samples obtained from such borings are highly disturbed. In some non-cohesive soils or soils having low cohesion, the walls of the boreholes will not stand unsupported. In such circumstances, a metal pipe is used as a casing to prevent the soil from caving in. When power is available, continuous-flight augers are probably the most common method used for advancing a borehole. The power for drilling is delivered by truck-or tractor-mounted drilling rigs. Boreholes up to about 60-70 m can easily be made by this method.Wash boring is another method of advancing boreholes. In this method, a casing about 2-3 m long is driven into the ground. The soil inside the casing is then removed by means of a chopping bit attached to a drilling rod. Water is forced through the drilling rod and exits at a very high velocity through the holes at the bottom of the chopping bit. The water and the chopped soil particles rise in the drill hole and overflow at the top of the casing through a T connection. The washwater is collected in a container. The casing can be extended with additional pieces as the borehole progresses; however, that is not required if the borehole will stay open and not cave in. Wash borings are rarely used now in the United States and other developed countries.

Rotary drilling is a procedure by which rapidly rotating drilling bits attached to the bottom of drilling rods cut and grind the soil and advance the borehole. There are several types of drilling bit. Rotary drilling can be used in sand, clay, and rocks (unless they are badly fissured). Water of drilling mud is forced down the drilling rods to the bits, and the return flow forces the cuttings to the surface. Boreholes with diameters of 50-203 mm (2-8 in.) can easily be made by this technique. The drilling mud is a slurry of water and bentonite. Generally, it is used when the soil that is encountered is likely to cave in. When soil samples are needed, the drilling rod is raised and the drilling bit is replaced by a sampler. With the environmental drilling applications, rotary drilling with air is becoming more common.Percussion drilling is an alternative method of advancing a borehole, particularly through hard soil and rock. A heavy drilling bit is raised and lowered to chop the hard soil. The chopped soil particles are brought up by the circulation of water. Percussion drilling may require casing.

Procedures for Sampling Soil

Two types of soil samples can be obtained during subsurface exploration: disturbed and undisturbed. Disturbed, but respresentative, samples can generally be used for the following types of laboratory test:

Grain-size analysis

Determination of liquid and plastic limits

Specific gravity of soil solids

Determination of organic content

Classification of soil

Disturbed soil samples, however, cannot be used for consolidation, hydraulic conductivity, or shear strength tests. Undisturbed soil samples must be obtained for these types of laboratory tests.

The degree of disturbance for a soil sample is usually expressed as:

Where:

AR= area ratio (ratio of disturbed area to total area of soil)

Do= outside diameter of the sampling tube

Di= inside diameter of the sampling tubeWhen the area ratio is 10% or less, the sample generally is considered to be undisturbed.

Split-Spoon Sampling (SPT)

Split-spoon samplers can be used in the field to obtain soil samples that are generally disturbed, but still representative. A section of a standard split-spoon sample is shown in figure.

Figure 8. Standard split-spoon sampler

The tool consists of a steel driving shoe, a steel tube that is split longitudinally in half, and a coupling at the top. The coupling connects the sampler to the drill rod. The standard split tube has an inside diameter of 34,93 mm ( in.) and an outside diameter of 50,8 mm (2 in.); however, samplers having inside and outside diameters up to 63,5 mm ( in.) and 76,2 mm (3 in.), respectively, are also available. When a borehole is extended to a predetermined depth, the drill tools are removed and the sampler is lowered to the bottom of the hole. The sampler is driven into the soil by hammer blows to the top of the drill rod. The standard weight of the hammer is 622,72 N (140 lb), and for each blow, the hammer drops a distance of 0,762 m (30 in.). The number of blows required for a spoon penetration of three 152,4 mm (6 in.) intervals are recorded. The number of blows required for the last two intervals are added to give the standard penetration number, N, at that depth. This number is generally referred to as the N value (American Society for Testing and Materials, 2001, Designation D-1586-99). The sampler is then withdrawn, and the shoe and coupling are removed. Finally, the soil sample recovered from the tube is placed in a glass bottle and transported to the laboratory. This field test is called the standard penetration test (SPT).The boring log shows refusal and the test is halted if :

50 blows are required for any 150-mm increment.

100 blows are obtained (to drive the required 300 mm).

10 successive blows produce no advance.For a standard split-spoon sampler,

When the area ratio is 10% or less, the sample generally is considered to be undisturbed. Hence, these samples are highly disturbed. Split-spoon samples generally are taken at intervals of about 1,53 m (5 ft). When the material encoutered in the field is sand (particularly fine sand below the water table), recovery of the sample by a split-spoon sampler may be difficult. In that case, a device such as a spring core catcher may have to be placed inside the split spoon.

Figure 9. Spring core catcher

At this juncture, it is important to point out that several factors contribute to the variation of the standard penetration number N at a given depth for similar soil profiles. Among these factors are the SPT hammer effeciency, borehole diameter, sampling method, and rod length factor (Skempton, 1986; Seed et al., 1985). The two most common types of SPT hammers used in the field are the safety hammer and donut hammer. They are commonly dropped by a rope with two wraps around a pulley.

Figure 10. (a) Safety hammer; (b) Donut hammer (after Seed et. Al., 1985)

Figure 11. The SPT sampler in place in the boring with hammer, rope and cathead (Kovacs, et al. 1981) Soil density or consistency description based on SPT blowcount values can be seen below (after AASHTO, 1988) :Table 9. Soil density/consistency based on SPT blowcount valuesCohesionless SoilsCohesive Soils

Relative DensityN-SPT values (blows/300 mm)ConsistencyN-SPT values (blows/300 mm)

Very loose

Loose

Medium dense

Dense

Very dense0 4

5 10

11 24

25 50

> 51Very soft

Soft

Medium stiff

Stiff

Very stiff

Hard

Very hard0 1

2 4

5 8

9 15

16 30

31 60

> 61

Cone Penetration Test (CPT)

The cone penetration test (CPT), originally known as the Dutch cone penetration test, is a versatile sounding method that can be used to determine the materials in a soil profile and estimate their engineering properties. The test is also called the static penetration test, and no borehole is necessary to perform it. In the original version, a 60( cone with a base area of 10 cm2 was pushed into the ground at a steady rate of about 20 mm/sec, and the resistance to penetration (called the point resistance) was measured.

The cone penetrometers in use at present measure (a) the cone resistance (qc) to penetration developed by the cone, which is equal to the vertical force applied to the cone, divided by its horizontally projected area; and (b) the frictional resistance (fc), which is the resistance measured by a sleeve located above the cone with the local soil surrounding it. The frictional resistance is equal to the vertical force applied to the sleeve, divided by its surface area actually, the sum of friction and adhesion.The original mechanical cone test is illustrated in Figure below with the step sequence as follows :

The cone system is stationary at position 1.

The cone is advanced by pushing an inner rod to extrude the cone tip and a short length of cone shaft. This action measures the tip resistance qc.

The outer shaft is now advanced to the cone base, and skin resistance is measured as the force necessary to advance the shaft fc.

Now the cone and sleeve are advanced in combination to obtain position 4 and to obtain a qtotal, which should be approximately the sum of the qc + fc just measured. The cone is now positioned for a new position 1.

Figure 12. Mechanical (or Dutch) cone, operations sequence, and tip resistance data Luas proyeksi ujung = 10 cm2 (ASTM D3411)

Luas selimut = 150 cm2 (ASTM D3411) atau ada juga 100 cm2Luas piston = 10 cm2 (ASTM D3411)

Berikut adalah penelusuran perhitungan bacaan pada sondir :

1. Bacaan I = tahanan ujung (R1)

qc x Aproyeksi = R1 x Apiston (qc, R1 = kg/cm2)

qc = R1 x (Apiston / Aproyeksi)2. Bacaan II = (tahanan ujung + tahanan selimut) (R2)

(qc x Aproyeksi) + (fs x Aselimut) = R2 x Apiston (qc, fs, R2 = kg/cm2)fs = {( R2 x Apiston)-( qc x Aproyeksi)}/ Aselimutfs = (R2- qc) x (Apiston or Aproyeksi / Aselimut)

dimana = (Apiston or Aproyeksi / Aselimut) faktor koreksi alatSeveral correlations that are useful in estimating the properties of soils encountered during an exploration program have been developed for the point resistance (qc) and the friction ratio (Fr) obtained from the cone penetration tests. The friction ratio is defined as :

Where:

fc= frictional resistance

qc = cone resistance

Fr= friction ratio

It may also be used to give an estimate of the soil sensitivity, St with the correlation being approximately :

Where:

St= sensitivityFr= friction ratio (in percent)

Clays may be classified as follows :

Shear Strength

The shear strength of a soil, defined in terms of effective stress, is

Where:

(= effective normal stress on plane of shearing

c= cohesion, or apparent cohesion

(= effective stress angle of friction

The equation above is referred to as the Mohr-Coulomb failure criterion. The value of c for sands and normally concolidated clays is equal to zero. For overconsolidated clays, c > 0.

For most day-to-day work, the shear strength parameters of a soil (i.e., c and () are determined by two standard laboratory tests: the direct shear test and the triaxial test.Direct Shear Test

Dry sand can be conveniently tested by direct shear tests. The sand is placed in a shear box that is split into two halves. First a normal load is applied to specimen. Then a shear force is applied to the top half of the shear box to cause failure in the sand. The normal and shear stresses at failure are:

and

Where:

A= area of the failure plan in soil - that is, the cross-sectional area of shear box R= applied shear forceSeveral tests of this type can be conducted by varying the normal load. The angle of friction of the sand can be determined by plotting a graph of ( against ( (= ( for dry sand), as shown in figure below, or :

For sands, the angle of friction usually ranges from 26( to 45(, increasing with the relative density of compaction. The approximate range of the relative density of compaction and the corresponding range of the angle of friction for various coarse-grained soils is shown in figure below:

Figure 13. Direct shear test in sand ; (a) schematic diagram of test equipment; (b) plot of test results to obtain the friction angle (A thin soil sample is placed in a shear box consisting of two parallel blocks. The lower block is fixed while the upper block is moved parallel to it in a horizontal direction. The soil fails by shearing along a plane assumed to be horizontal.

This test is relatively easy to perform. Consolidated-drained tests can be performed on soils of low permeability in a short period of time as compared to the triaxial test. However, the stress, strain, and drainage conditions during shear are not as accurately understood or controlled as in the triaxial test.

Triaxial Test

Two fundamentally different approaches to the solution of stability problems in geotechnical engineering :

1. The total stress approach

In the total stress approach, we allow no drainage to take place during the shear test, and we make the assumption, admittedly a big one, that the pore water pressure and therefore the effective stresses in the test specimen are identical to those in the field. The method of stability analysis is called the total stress analysis, and it utilizes the total or the undrained shear strength (f, of the soil. The undrained strength can be determined by either laboratory or field tests. If field tests such as the vane shear, Dutch cone penetrometer, or pressuremeter test are used, then they must be conducted rapidly enough so that undrained conditions prevail in situ.

2. The effective stress approach

The second approach to calculate the stability of foundations, embankments, slopes, etc., uses the shear strength in terms of effective stresses. In this approach, we have to measure or estimate the excess hydrostatic pressure, both in the laboratory and in the field. Then, if we know or can estimate the initial and applied total stresses, we may calculate the effective stresses acting in the soil. Since we believe that shear strength and stress-deformation behavior of soils is really controlled or determined by the effective stresses, this second approach is philisophically more satisfying. But, it does have its practical problems. For example, estimating or measuring the pore pressures, especially in the field, is not easy to do. The method of stability is called the effective stress analysis, and it utilizes the drained shear strength or the shear strength in terms of effective stresses. The drained shear strength is ordinarily only determined by laboratory tests.Triaxial tests can be conducted on sands and clays. Essentially, the test consists of placing a soil specimen confined by a rubber membrane into a lucite chamber and then applying an all-arround confining pressure ((3) to the specimen by means of the chamber fluid (generally, water or glycerin). An added stress ((() can also be applied to the specimen in the axial direction to cause failure ((( = ((f at failure). Drainage from the specimen can be allowed or stopped, depending on the condition being tested. For clays, three main types of tests can be conducted with triaxial equipment:

Figure 14. Schematic diagram of triaxial test equipment

Figure 15. Sequence of stress application in triaxial test1. Consolidated Drained test (CD test)

Step 1. Apply chamber pressure (3. Allow complete drainage, so that the pore water pressure (u = u0) developed is zero.

Step 2. Apply deviator stess (( slowly. Allow drainage, so that the pore water pressure (u = ud) developed through the application of (( is zero. At failure, (( = ((f ; the total pore water pressure uf = u0 + ud = 0.

Figure 16. Stress conditions in the consolidated-drained (CD) axial compression triaxial testSo for consolidated-drained tests, at failure,

Major principal effective stress = (3 + ((f = (1 = (1

Minor principal effective stress = (3 = (3

Changing (3 allows several tests of this type to be conducted on various clay specimens. The shear strength parameters (c and () can now be determined by plotting Mohrs circle at failure, and drawing a common tangent to the Mohrs circles. This is the Mohr-Coulomb failure envelope. (Note: For normally consolidated clay, c 0 ; For overconsolidated clay, c > 0) At failure,

Figure 17. Consolidated-drained testThe envelope for a normally consolidated clay is shown below. Even though only one Mohr circle (representing the stress conditions at failure) is shown, the results of three or more CD tests on identical specimens at different consolidation pressures would ordinarily be required to plot the complete Mohr failure envelope. If the consolidation stress range is large or the specimens do not have exactly the same initial water content, density, and stress history, then the three failure circles will not exactly define a straight line, and an average best-fit line by eye is drawn. The slope of the line determines the Mohr-Coulomb strength parameter (, of course, in terms of effective stresses. When the failure envelope is extrapolated to the shear axis, it will show a surprisingly small intercept. Thus it is usually assumed that the c parameter for normally consolidated non-cemented clays is essentially zero for all practical purposes.

Figure 18. Mohr failure envelope for a normally consolidated clay in drained shearFor overconsolidated clays the c parameter is greater than zero, as indicated by figures below. The overconsolidated portion of the strength envelope (DEC) lies above the normally consolidated envelope (ABCF). This portion (DEC) of the Mohr failure envelope is called the preconsolidated hump. The explanation for this behavior is shown in the e versus ( curve of figure below. Let us assume that we begin consolidation of a sedimentary clay at a very high water content and high void ratio. As we continue to increase the vertical stress we reach point A on the virgin compression curve and conduct a CD triaxial test. The strength of the sample consolidated to point A on the virgin curve would correspond to point A on the normally consolidated Mohr failure envelope in figure below. If we consolidate and test another otherwise identical specimen which is loaded to point B, then we would obtain the strength, again normally consolidated, at point B on the failure envelope in figure below. If we repeat the process to point C ((p, the preconsolidation stress), then rebound the specimen to point D, then reload it to point E and shear, we would obtain the strength shown at point E is greater than specimen B, even though they are tested at exactly the same effective consolidation stresses. The reason for the greater strength of E than B is suggested by the fact that E is at a lower water content, has a lower void ratio, and thus is denser than B, as shown in figure below. If another specimen were loaded to C, rebounded to D, reloaded back past E and C and on to F, it would have the strength as shown in figure at point F. Note that it is now back on the virgin compression curve and the normally consolidated failure envelope. The effects of the rebounding and reconsolidation have been in effect erased by the increased loading to point F. Once the soil has been loaded well past the preconsolidation pressure (p, it no longer remembers its stress history.

Figure 19. (a) Compression curve; (b) Mohr failure envelope (DEC) for an overconsolidated clayIn the CD test, complete consolidation of the test specimen is permitted under the confining pressure and drainage is permitted during shear. The rate of strain is controlled to prevent the build-up of pore pressure in the specimen. A minimum of three tests are required for c and ( determination. CD tests are generally performed on well draining soils. For slow draining soils, several weeks may be required to perform a CD test.Typical stress-strain curves and volume change versus strain curves for a remolded or compacted clay are shown below. Even though the two samples were tested at the same confining pressure, the overconsolidated specimen has a greater strength than the normally consolidated clay. Note also that it has a higher modulus and that failure [the maximum ((, which for the triaxial test is equal to ((1 - (3)f] occurs at a much lower strain that for the normally consolidated specimen. The overconsolidated clay expands during shear while the normally consolidated clay compresses or consolidates during shear.

Figure 20. Typical stress-strain and volume change versus strain curves for CD axial compression tests at the same effective confining stressAverage values of ( for undisturbed clays range from around 20 for normally consolidated highly plastic up to 30 or more for silty and sandy clays. The value of ( for compacted clays is typically 25( or 30 and occasionally as high as 35. The value of c for normally consolidated non-cemented clays is very small and can be neglected for practical work. If the soil is overconsolidated, then ( would be less, and the c intercept greater than for the normally consolidated part of the failure envelope. According to Ladd (1971b), for natural overconsolidated non-cemented clays with a preconsolidation stress of less than 500 to 1000 kPa, c will probably be less than 5 to 10 kPa at low stresses. For compacted clays at low stresses, c will be much greater due to the prestress caused by compaction. For stability analyses, the Mohr-Coulomb effective stress parameters ( and c are determined over the range of effective normal stresses likely to be encountered in the field.

Empirical correlations between ( and the plasticity index for normally consolidated clays are shown below:

Figure 21. Emperical correlation between ( and PI from triaxial compression tests on normally consolidated undisturbed clays (after U.S. Navy, 1971, and Ladd, et al., 1977)Where do we use the strengths determined from the CD test? As mentioned previously, the limiting drainage conditions modeled in the triaxial test refer to real field situations. CD conditions are the most critical for the long-term steady seepage case for embankment dams and the long-term stability of excavations or slopes in both soft and stiff clays. The examples of CD analysis can be seen below:

Figure 22. Some examples of CD analyses for clays (after Ladd, 1971b)2. Consolidated Undrained test (CU test)

Step 1. Apply chamber pressure (3. Allow complete drainage, so that the pore water pressure (u = u0) developed is zero.Step 2. Apply a deviator stress ((. Do not allow drainage, so that the pore water pressure u = ud 0. At failure, (( = ((f ; the pore water pressure uf = u0 + ud = 0 + u d(f).

Note that the excess pore water pressure (u developed during shear can either be positive (that is, increase) or negative (that is, decrease). This happens because the sample tries to either contract or expand during shear. Remember, we are not allowing any volume change (an undrained test) and therefore no water can flow in or out of the specimen during shear. Because volume changes are prevanted, the tendency towards volume change induces a pressure in the pore water. If the specimen tends to contract or consolidate during shear, then the induced pore water pressure is positive. It wants to contract and squeeze water out of the pores, but cannot; thus the induced pore water pressure is positive. Positive pore pressures occur in normally consolidated clays. If the specimen tends to expand or swell during shear, the induced pore water pressure is negative. It wants to expand and draw water into the pores, but cannot; thus the pore water pressure decreases and may even go negative (that is, below zero gage pressure). Negative pore pressures occur in overconsolidated clays.Hence, at failure,

Major principal total stress = (3 + ((f = (1

Minor principal total stress = (3

Major principal effective stress = ((3 + ((f) uf = (1

Minor principal effective stress = (3 uf = (3

Figure 23. Conditions in specimen during a consolidated-undrained axial compression (CU) testChanging (3 permits multiple tests of this type to be conducted on several soil specimens. The total stress Mohrs circles at failure can now be plotted, and then a common tangent can be drawn to define the failure envelope. This total stress failure envelope is defined by the equation

Where c and ( are the consolidated-undrained cohesion and angle of friction, respectively (Note: c 0 for normally consolidated clays)Similarly, effective stress Mohrs circles at failure can be drawn to determine the effective stress failure envelope, which satisfy the relation .

Figure 24. Consolidated-undrained testIn the CU test, complete consolidation of the test specimen is permitted under the confining pressure, but no drainage is permitted during shear. A minimum of three tests is required to define strength parameters c and (, through four test specimens are preferable with one serving as a check. Specimens must as a general rule be completely saturated before application of the deviator stress. Full saturation is achieved by back pressure. When a back pressure is applied to a sample, the cell pressure must also be increased by an amount equal to the back pressure so that the effective consolidation stresses will remain the same. Since the effective stress in the specimen does not change, the strength of the specimen is not supposed to be changed by the use of back pressure. In practice this may not be exactly true, but the advantage of having 100% saturation for accurate measurement of induced pore water pressures far outweighs any disadvantages if the use of back pressure.

Typical stress-strain, (u, and (1/(3 curves for CU tests are shown below, for both normally and overconsolidated clays. Also shown for comparison is a stress-strain curve for an overconsolidated clay at low effective consolidation stress. Note the peak, then the drop-off of stress as strain increases (work-softening material). The pore pressure versus strain curves illustrate what happens to the pore pressures during shear. The normally consolidated specimen develops positive pore pressure. In the overconsolidated specimen, after a slight initial increase, the pore pressure goes negative in this case, negative with respect to the back pressure u0. Another quantity that is useful for analyzing test results is the principal (effective) stress ratio (1/(3. Note how this ratio peaks early, just like the stress difference curve, for the overconsolidated clay. Similar test specimens having similar behavior on an effective stress basis will have similarly shaped (1/(3 curves. They are simply a way of normalizing the stress behavior with repect to the effective minor principal stress during the test. Sometimes, too, the maximum of this ratio is used as a criterion of failure. However, in this text we will continue to assume failure occurs at the maximum principal stress difference (compressive strength).

Figure 25. Typical (-(, (u, and (1/(3 curves for normally and overconsolidated clays in undrained shear (CU test)Since we can get both the total and effective stress circles at failure for a CU test when we measure the induced pore water pressures, it is possible to define the Mohr failure envelopes in terms of both total and effective stresses from a series of triaxial tests conducted over a range of stresses, as illustrated in figure below for a normally consolidated clay.

Figure 26. Mohr circles at failure and Mohr failure envelopes for total (T) and effective (E) stresses for a normally consolidated clayNote that the effective stress circle is displaced to the left, towards the origin, for the normally consolidated case, because the specimens develop positive pore pressure during shear and ( = ( - (u. Note that both circles have the same diameter because of our definition of failure at maximum ((1 - (3) = ((1 - (3). Once the two failure envelopes are drawn, the Mohr-Coulomb strength parameters are readily definable in terms of both total (c, ( or sometimes cT, (T) and effective stresses (c, (). Again, as with the CD test, the envelope for normally consolidated clay passes essentially through the origin, and thus for practical purposes c can be taken to be zero, which is also true for the total stress c parameter. Note that (T is less than (, and often it is about one-half of (.Things are different if the clay is overconsolidated. Since an overconsolidated specimen tends to expand during shear, the pore water pressure decreases or even goes negative, as shown in previous figure. Because (3f = (3f (-(uf) or (1f = (1f (-(uf), the effective stresses are greater than the total stresses, and the effective stress circle at failure is shifted to the right of the total stress circle as shown in figure below. The shift of the effective stress circle at failure to the right sometimes means that the ( is less than (T. Typically, the complete Mohr failure envelopes are determined by tests on several specimens consolidated over the working stress range of the field problem.

Figure 27. Mohr circles at failure and Mohr failure envelopes for total (T) and effective (E) stresses for an overconsolidated clayFigure below shows the Mohr failure envelopes over a wide range of stresses spanning the preconsolidation stress. Thus some of the specimens are overconsolidated and others are normally consolidated. You should note that the break in the total stress envelope (point z) occurs roughly about twice the (p for typical clays (Hirschfeld, 1963). The two sets of Mohr circles at failure shown in figure below correspond to the two tests for the normally consolidated specimen and the specimen overconsolidated at low (hc.

Figure 28. Mohr failure envelopes over a range of stresses spanning the preconsolidation stress (pNote :

Pore water pressure is measured during the CU test, thus permitting determination of the effective stress parameters c and (. In the absence of pore pressure measurements CU tests can provide only total stress values c and (.

3. Unconsolidated Undrained test (UU test)

Step 1. Apply chamber pressure (3. Do not allow drainage, so that the pore water pressure (u = u0) developed through the application of (3 is not zero.Step 2. Apply a deviator stress ((. Do not allow drainage (u = ud 0). At failure, (( = ((f ; the pore water pressure uf = u0 + ud(f)For unconsolidated-undrained triaxial tests,

Major principal total stress = (3 + ((f = (1

Minor principal total stress = (3

Figure 29. Conditions in the specimen during the unconsolidated-undrained (UU) axial compression testThe total stress Mohrs circle at failure can now be drawn. For saturated clays, the value of (1 - (3 = ((f is a constant, irrespective of the chamber confining pressure (3. The tangent to these Mohrs circles will be a horizontal line, called the ( = 0 condition. The shear stress for this condition is

Where cu = undrained cohesion (or undrained shear strength)

Figure 30. Unconsolidated-undrained testThe pore pressure developed in the soil specimen during the unconsolidated-undrained triaxial test is

The pore pressure ua is the contribution of the hydrostatic chamber pressure (3. Hence,

Where B = Skemptons pore pressure parameter

Similarly, the pore parameter ud is the result of the added axial stress ((, so

Where A = Skemptons pore pressure parameter

However,

Combining the equations above gives

The pore water pressure parameter B in soft saturated soils is unity, so

The value of the pore water pressure parameter A at failure will vary with the type of soil. Following is a general range of the values of A at failure for various types of clayey soil encountered in nature:

Table 10. Values of A at failure for various types of clayey soilType of SoilA at failure

Sandy clays0,5 0,7

Normally consolidated clays0,5 1,0

Overconsolidated clays-0,5 0,0

Unconfined Compression Test

The unconfined compression test is a special type of unconsolidated-undrained triaxial test in which the confining pressure (3 = 0. In this test, an axial stress (( is applied to the specimen to cause failure (i.e., (( = ((f). The corresponding Mohrs circle is shown in figure. Note that, for this case,

Figure 31. Soil specimen

Figure 32. Mohrs circle for the unconfined compression testMajor principal total stress = ((f = qu

Minor principal total stress = 0

The axial stress at failure, ((f = qu, is generally referred to as the unconfined compression strength. The shear strength of saturated clays under this condition (( = 0) is

The unconfined compression strength can be used as an indicator of the consistency of clays.

Unconfined compression tests are sometimes conducted on unsaturated soils. With the void ratio of soil specimen remaining constant, the unconfined compression strength rapidly decreases with the degree of saturation.

Figure 33. Variation of qu with the degree of saturationSensitivity

For many naturally deposited clay soils, the unconfined compression strength is much less when the soils are tested after remolding without any change in the moisture content. This property of clay soil is called sensitivity. The degree of sensitivity is the ratio of the unconfined compression strength in an undisturbed state to that in a remolded state, or

The sensitivity ratio of most clays from about 1 to 8; however, highly flocculent marine clay deposits may have sensitivity ratios ranging from about 10 to 80. Some clays turn to viscous liquids upon remolding, and these clays are referred to as quick clays. The loss of strength of clay soils from remolding is caused primarily by the destruction of the clay particle structure that was developed during the original process of sedimentation.

Based on Holtz and Kovacs, 1981:

St 4

low sensitivity

4 < St 8

medium sensitivity

8 < St 16

high sensitivity

St > 16

quick sensitivity

Note: when testing a remolded soil specimen, it is important to retain the same water content of the undisturb soil. If the soil specimen bleeds water during this process, then the sensitivity cannot be determined for the soil.An unsual feature of highly sensitive or quick clays is that the insitu water content is often greater than the liquid limit (liqudity index greater than one).

Sensitive clays have unstable bonds between particles. As long as these unstable bonds are not broken, the clay can support a heavy load. But once remolded, the bonding is destroyed and the shear strength is substantially reduced.

Vane Shear Test

The vane shear test (ASTM D-2573) may be used during the drilling operation to determine the in situ undrained shear strength (cu) of clay soils particularly soft clays. The vane shear apparatus consists of four blades on the end of a rod, as shown in Figure below. The height, H, of the vane is twice the diameter, D. The vane can be either rectangular of tapered. The dimensions of vanes used in the field are given in Table below. The vanes of the apparatus are pushed into the soil at the bottom of a borehole without disturbing the soil appreciably. Torque is applied at the top of the rod to rotate the vanes at a standard rate of 0,1(/sec (6(/min). This rotation will induce failure in a soil of cylindrical shape surrounding the vanes. The maximum torque, T, applied to cause failure is measured. Note thatT = f (cu, H, and D)or

Where:T = torque (N.m)

cu = undrained shear strength (kN.m2)

K = a constant with a magnitude depending on the dimension and shape of vaneThe constant

Where:

D = diameter of vane (cm)

H = measured height of vane (cm)

K = a constant with a magnitude depending on the dimension and shape of vane

Figure 34. Geometry of Field Vane (after ASTM, 2001)Table 11. Recommended Dimensions of Field Vanesa (After ASTM, 2001)Casing SizeDiamter, D

mm (in.)Height, H

mm (in.)Thickness of Blade

mm (in.)Diameter of Rod

mm (in.)

AX38,1

76,2

1,6

12,7

BX50,8

101,6

1,6

12,7

NX63.5

127,0

3,2

12,7

4 in (101,6 mm)b92,1

184,1

3,2

12,7

aThe selection of a vane size is directly related to the consistency of the soil being tested; that is, the softer the soil, the larger the vane diameter should be.bInside diameter.Field vane shear tests are moderately rapid and economical and are used extensively in field soil-exploration programs. The test gives good results in soft and medium-stiff clays and gives excellent results in determining the properties of sensitive clays.Sources of significant error in the field vane shear test are poor calibration of torque measurement and damaged vanes. Other errors may be introduced if the rate of rotation of the vane is not properly controlled.

For actual design purposes, the undrained shear strength values obtained from field vane shear tests [cu(VST)] are too high, and it is recommended that they be corrected according to the equation

Where:cu = undrained shear strength

= correction factor

Several correlations have been given previously for the correction factor (; some more are as follows:

Bjerrum (1972):

Morris and Williams (1994):

(for PI > 5%)

(LL in %)Aas et al. (1986):

(see figure below)

Figure 35. Variation of ( with cu(VST)/(0Atterberg LimitsWhen a clayey soil is mixed with an excessive amount of water, it may flow like a semiliquid. If the soil is gradually dried, it will behave like a plastic, semisolid, or solid material, depending on its moisture content. The moisture content, in percent, at which the soil changes from a liquid to a plastic state is defined as the liquid limit (LL). Similarly, the moisture content, in percent, at which the soil changes from a plastic to a semisolid state and from a semisolid to a solid state is defined as the plastic limit (PL) and the shrinkage limit (SL), repectively. These limits are referred to as Atterberg Limits:

The liquid limit of a soil is determined by Casagrandes liquid device (ASTM Test Designation D-4318) and is defined as the moisture content at which a groove closure of 12.7 mm (1/2 in.) occurs at 25 blows.

The plastic limit is defined as the moisture content at which the soil crumbles when rolled into a thread of 3.18 mm (1/8 in.) in diameter (ASTM Test Designation D-4318).

The shrinkage limit is defined as the moisture content at which the soil does not undergo any further change in volume with loss of moisture (ASTM Test Designation D-4318).

The difference between the liquid limit and the plastic limit of a soil is defined as the plasticity index (PI), or :

However, Atterberg limits for the different soils will vary considerably, depending on the origin of the soil and the nature and amount of clay minerals in it.

Figure 36. Definition of Atterberg limitsThe liquid and plastic limit values, together with wN (natural water content), are useful in predicting whether a cohesive soil mass is preconsolidated. Since an overconsolidated soil is more dense, the void ratio is smaller than in the soil remolded for the Atterberg limit tests. If the soil is located below the groundwater table (GWT) where it is saturated, one would therefore expect that smaller void ratios would have less water space and the wN value would be smaller. From this we might deduce the following:

If wN is close to wL, soil is normally consolidated

If wN is close to wP, soil is some- to heavily overconsolidated

If wN is intermediate, soil is somewhat overconsolidated

If wN is greater than wL, soil is on verge of being a viscous liquid

Although the foregoing gives a qualitative indication of overconsolidation, other methods must be used if a quantitative value of OCR is required.

Overconsolidation Ratio (OCR)

The degree of overconsolidation may be expressed numerically as the overconsolidation ratio (OCR) which is defined as follows:

Where:

OCR= overconsolidation ratio

(c= pre-consolidation ratio

(v= vertical effective stress

The OCR of a normally consolidated soil is equal to 1; in a lightly overconsolidated soil, it is typically between 1 and 3; a heavily overconsolidated soil may have an OCR as high as 8. Pre-consolidation stress can be determined by consolidation test.

The following empirical relationship (U.S. Navy, 1982a) can be useful when estimating it or when checking test result for reasonableness:

Where:(c= pre-consolidation ratio

cu= undrained shear strength

PI= Plasticity IndexSynthesis of Field and Laboratory Data

Investigation and testing programs often generate large amounts of information that can be difficult to sort through and synthesize. Real soil profiles are nearly always very complex, so the borings will not correlate and the test results will often vary significantly. Therefore, we must develop a simplified soil profile before proceeding with the analysis. In many cases, the simplified profile is the best defined in terms of a one-dimensional function of soil type and engineering properties vs. depth; an idealized boring log. However, when the soil profile varies significantly across the site, one or more vertical cross-sections may be in ordered,

The development of these simplified profiles requires a great deal of engineering judgment along with interpolation and extrapolation of the data. It is important to have a feel for the approximate magnitude of the many uncertainties in this process and reflect them in an appropriate degree of conservatism. This judgment comes primarily with experience combined with a through understanding of the field and laboratory methodologies.

Organic Soil

Organic soils are usually found in low-lying areas where the water table is near or above the ground surface. The presence of a high water table helps in the growth of aquatic plants that, when decomposed, form organic soil. This type of soil deposit is usually encountered in coastal areas and in glaciated regions. Organic soils show the following characteristics:

Their natural moisture content may range from 200% to 300%

They are highly compresible

Laboratory tests have shown that, under loads, a large amount of settlement is derived from secondary consolidation

Expansive Soil

When geotechnical engineers refer to expansive soils, we usually are thinking about clays or sedimentary rocks derived from clays, and the volume changes that occur as a result of changes in moisture content. Clays are fundamentally very different from gravels, sands, and silts. All of the later consist of relatively inert bulky particles and their engineering properties depend primarily on the size, shape, and texture of these particles. In contrast, clays are made of very small particles that are usually plate-shaped. The engineering properties of clays are strongly influenced by the very small size and large surface area of these particles and their inherent electrical charges.

Several different clay minerals occur in nature, the differences being defined by their chemical makeup and structural configuration. Three of the most common clay minerals are kaolinite, illite, and montmorillonite (part of the smectite group). The different chemical compositions and crystalline structures of these minerals give each a different susceptibility to swelling, as shown in table below.

Table 12. Swell Potential Of Pure Clay Minerals (Budge et al., 1964)Sucharge LoadSwell Potential (%)

(lb/ft2)(kPa)KaoliniteIlliteMontmorillonite

2009.6Negligible3501500

40019.1Negligible150350

Swelling occurs when water infiltrates between and within the clay particles, causing them to separate. Kaolinite is essentially nonexpansive because of the presence of strong hydrogen bonds that hold the individual clay particles together. Illite contains weaker potassium bonds that allow limited expansion, and montmorillonite particles are only weakly linked. Thus, water can easily flow into montmorillonite clays and separate the particles. Field observations have confirmed that the greatest problems occur in soils with a high montmorillonite content.

Several other forces also act on clay particles, including the following:

Surface tension in the menisci of water contained between the particles (tends to pull the particles together, compressing the soil).

Osmotic pressures (tend to bring water in, thus pressing the particles further apart and expanding the soil).

Pressures in entrapped air bubles (tend to compress the soil).

Effective stresses due to external loads (tend to compress the soil).

London-Van Der Waals intermolecular forces (tend to compress the soil).

Expansive clays swell or shrink in response to changes in these forces. For example, consider the effects of changes in surface tension and osmotic forces by imagining a montmorillonite clay that is initially saturated, as shown if figure (a). If this soil dries, the remaining moisture congregates near the particle interfaces, forming menisci, as shown in figure (b), and the resulting surface tension forces pull the particles closer together causing soil to shrink. We could compare the soil in this stafe to a compressed spring: both would expand if it were not for forces keeping them compressed. The soil in figure (b) has a great affinity for water and will draw in available water using osmosis. We would say that it has a very high soil suction at this stage. If water becomes available, the suction will draw it into spaces between the particles and the soil will swell, as shown in figure (c).

Internal Friction (()

Internal friction is used to analyze the bearing capacity of a foundation on Sand layer. This chart shows relationships between angle of friction and (N1)60.

Figure 37. Relationship Between Angle of Friction and N-Value for Sandy Soil (K. Terzaghi)

Based on the chart above, we propose to use Dunhams (1954) equation as:

Where:

(= angle of friction ( 0)

(N1)60= N-SPT value corrected for field procedures and overburden stressDeMello (1971) suggested a correlation between SPT data results and the effective friction angle of uncemented sands, (, as shown in Figure below. This correlation should be used only at depths greater than about 2 m.

Figure 38. Empirical correlation between N60 and ( for uncemented sands (DeMello, 1971)

The CPT results also have been correlated with shear strength parameters, especially in sands. Figure below presents Robertson and Campanellas 1983 correlation for uncemented, normally consolidated quartz sands. For overconsolidated sands, subtract 10 to 20 from the effective friction angle obtained from this figure.

Figure 39. Relationship Between CPT Results, Overburden Stress and Effective Friction Angle for Uncemented, Normally Consolidated Quartz Sand (Robertson and Campanella, 1983)

On the basis of experimental results, Robertson and Campanella (1983) suggested the variation of ( for normally consolidated quartz sand. The figure above can also be expressed into a relationship as (Kulhawy and Mayne, 1990)

(= effective angle of friction ( 0)qc= cone resistance(0 = effective vertical stressCohesion (cu)Cohesion is one of the soil properties for clay that can be used to analyze the bearing capacity of a foundation. The chart shows the relationships between undrained cohesion and corrected N-SPT value :

Figure 40. Relationship Between Cohesion with N-Value for Cohesive Soil (K. Terzaghi)

The chart above is used to determine the undrained shear strength for cohesive soils and we proposed to use:

Where:

cu= undrained shear strength

N60 = N-SPT value corrected for field procedures

The literature contains many correlations between the standard penetration number and the undrained shear strength of clay, cu. On the basis of results of undrained triaxial tests conducted on insensitive clays, Stroud (1974) suggested that

Where:

cu= undrained shear strength

N60 = N-SPT value corrected for field procedures K= constant = 3,5 6,5Hara et al. (1971) also suggested that

As in the case of standard penetration tests, several correlations have been developed between qc and other soil properties. According to Mayne and Kemper (1988), in clayey soil the undrained cohesion cu, can be correlated via the equation

Where:

cu= undrained shear strengthqc= cone resistance(0 = total vertical stress

NK = 15 for an electric cone

= 20 for a mechanical coneValues of the undrained shear strength cu corresponding to various degrees of consistency are as follows (Terzaghi & Peck, 1967 and ASTM D2488-90) :

Very Soft: cu < 12 kPa. The clay is easily penetrated several centimeters by the thumb. The clay oozes out between the fingers when squeezed in the hand.

Soft: 12 kPa cu < 25 kPa. The clay is easily penetrated 2 to 3 cm by the thumb. The clay can be molded by slight finger pressure.

Medium: 25 kPa cu < 50 kPa. The clay can be penetrated about 1 cm by the thumb with moderate effort. The clay can be molded by strong finger pressure. Stiff: 50 kPa cu < 100 kPa. The clay can be indented about 0.5 cm by the thumb with great effort.

Very Stiff: 100 kPa cu < 200 kPa. The clay cannot be indented by the thumb, but can be readily indented with the thumbnail.

Hard: cu 200 kPa. With great difficulity, the clay can only be indented with the thumbnail.

Relative Density (Dr)The following approximate relationship between CPT results and the relative density of sands (Adapted from Kulhawy and Mayne, 1990) :

Where:

qc= cone resistanceQc= compressibility factor

= 0.91 for highly compressible sands

= 1.00 for moderately compressible sands

= 1.09 for slightly compressible sands

OCR= overconsolidation ratio(z = vertical effective stress

A relationship between consistency of sands and gravels and relative density is shown in Table below.

Table 13. Consistency of coarse-grained soils various relative densities (Lambe and Whitman, 1969)Relative Density, Dr (%)Classification

0 15Very Loose

15 - 35Loose

35 65Medium Dense

65 85Dense

85 100Very Dense

qc vs N-SPT ValueSince the SPT and CPT are the two most common in-situ tests, it often is useful to convert results from one to the other. The ratio qc/N60 as a function of the mean grain size, D50, is shown in Figure below. Note that N60 does not include an overburden correction.

Figure 41. Correlation between qc/N60 and the mean grain size, D50 (Kulhawy and Mayne, 1990)Be cautious about converting CPT data to equivalent N values, and then using SPT based analysis methods. This technique compounds the uncertainties because it uses two correlations one to convert to N, and then another to compute the desired quantity.Correlation with Soil ClassificationBecause the CPT does not recover any soil samples, it is not a substitute for conventional exploratory borings. However, it is possible to obtain an approximate soil classification using the correlation shown in Figure below:

Figure 42. Classification of Soil Based On CPT Test Results (Robertson and Campanella, 1983)

Design N-SPT ValueEarly recommendations were to use the smallest N value in the boring or an average of all of the values for the particular stratum. Current practice is to use an average N but in the zone of major stressing. For example, for a spread footing the zone of interest is from about one-half the footing width B above the estimated base location to a depth of about 2B below. Weighted averaging using depth increment multiplied by N may be preferable to an ordinary arithmetic average;

that is,

and not

For pile foundations there may be merit in the simple averaging of blow count N for any stratum unless it is very thick thick being a relative term. Here it may be better to subdivide the thick stratum into several strata and average the N count for each subdivision.

Correction for N-SPT ValueDeep tests in a uniform soil deposit will have higher N values than shallow tests in the same soil, so the overburden correction adjusts the measure N values to what they would have been if the vertical effective stress, (v, was 100 kPa. In granular soils, the value of N is affected by the effective overburden pressure, (0. For that reason, the value of N60 obtained from field exploration under different effective overburden pressure should be changed to correspond to a standard value of (0.So that, the corrected value (Liao and Whitman, 1985), (N1)60 is:

Where:

(N1)60= N-SPT value corrected for field procedures and overburden stress

CN = overburden correction factor (see figure below)(0= vertical effective stress in ton/ft2, which is based on water table during SPT testing. If fill is paced after SPT testing, fill does not affect (0

Figure 43. SPT overburden stress correction factor, CN (Liao and Whitman, 1986)Liao and Whitmans relationship (1986):

Skemptons relationship (1986):

Seed et al.s relationship (1975):

Peck et al.s relationship (1974):

We can improve the raw SPT data by applying certain correction factors. The variations in testing procedures may be at least partially compensated by converting the measured N to N60 as follows (Skempton, 1986):

Em= hammer efficiency

Cb= borehole diameter correction

Cs= sampling method correction

Cr= rod length correction

N= N-SPT value from field test

N60 = N-SPT value corrected for field procedures

(v

= vertical effective stress at the test locationPa

= reference stress = 100 kN/m2Table 14. SPT Hammer Effeciences (Clayton, 1990)CountryHammer TypeHammer Release MechanismHammer Effeciency, Em

ArgentinaDonutCathead0.45

BrazilPin weightHand dropped0.72

ChinaAutomaticTrip0.60

DonutHand dropped0.55

DonutCathead0.50

ColombiaDonutCathead0.50

JapanDonutTombi trigger0.78 0.85

DonutCathead 2 turns + special release0.65 0.67

UKAutomaticTrip0.73

USSafety2 turns on cathead0.55 0.60

Donut2 turns on cathead 0.45

VenezuelaDonutCathead0.43

Figure 44. Type of SPT hammersTable 15. Borehole, Sampler and Rod Correction Factors (Skempton, 1986)FactorEquipment VariablesValue

Borehole diameter factor, CB65 115 mm (2.5 4.5 in)1.00

150 mm (6 in)1.05

200 mm (8 in)1.15

Sampling method factor, CSStandard sampler1.00

Sampler without liner (not recommended)1.20

Rod length factor, CR3 4 m (10 13 ft)0.75

4 6 m (13 20 ft)0.85

6 10 m (20 30 ft)0.95

> 10 m (> 30 ft)1.00

Although Liao and Whitman did not place any limits in this correction, it is probably best to keep (N1)60 2.N60. This limit avoids excessively high (N1)60 values at shallow depths.The use of correction factors is often a confusing issue. Corrections for field procedures are always appropriate, but the overburden correction may or may not be appropriate depending on the procedures used by those who developed the analysis method under consideration. The N-SPT value, as well as many other test results, is only an index of soil behavior. It does not directly measure any of the conventional engineering properties of soil and is useful only when appropriate correlations are available. Many such correlation exist, all of which were obtained empirically. Be especially cautious when using correlations between SPT results and engineering properties of clays because these functions are especially crude. In general, the SPT should be used only in sandy soils.Adhesion Factor (( )Recommended values of ( for drilled shafts in clay (After Reese and ONeill, 1986):Table 16. Adhesion Value for Drilled Shaft in Clay (After Reese and ONeill,1986)Location Along Drilled ShaftUndrained Shear Strength (cu) ;

1 tsf = 95,76 kPaValue of (

From Ground surface to depth along drilled shaft of 5 ft--0

Bottom 1 diameter of the drilled shaft or 1 stem diameter above the top of the bell--0

All other points along the sides of the drilled shaft< 2 tsf< 191,52 kPa0,55

2 3 tsf191,52 287,28 kPa 0,49

3 4 tsf287,28 383,04 kPa0,42

4 5 tsf383,04 478,80 kPa0,38

5 6 tsf478,80 574,56 kPa0,35

6 7 tsf574,56 670,32 kPa0,33

7 8 tsf670,32 766,08 kPa0,32

8 9 tsf766,08 861,85 kPa0,31

> 9 tsf> 861,85 kPaTreat as Rock

Adhesion factor (() as function of undrained shear strength (su) for drilled shafts foundation in clay is presented below:

Figure 45. Adhesion Factor (() vs Undrained Shear Strength (su) for drilled shaftRecommended values of ( for pile in clay (API): For cu < 25 kPa

For 25 kPa < cu < 75 kPa

For cu > 75 kPa

Figure 46. Adhesion Factor (() vs Undrained Shear Strength (su) for driven pile (API)All of the factors presented above are for insensitive clays (St < 4). In sensitive clays, full-scale static load tests, special lab tests, or some other method of verification are appropriate (ONeill and Reese, 1999).

ONeill and Reese (1989) also ignore the skin friction resistance in the upper 5 ft (1,5 m) of the shaft and along the bottom one diameter of straight shaft because of interaction with the end bearing.Soil Modulus Subgrade Reaction and Soil StrainAccording to Reese (1987), modulus subgrade reaction and soil strain can be determained from the table below:

For ClaysTable 17. Modulus Subgrade Reaction and Soil Strain Value for Clay (after Reese, 1987)Consistencycu (kPa)k (kPa/m)(50

Soft12 2481400,02

Medium24 48271500,01

Stiff48 961360000,007

Very Stiff96 1922710000,005

Hard192 - 3835430000,004

For Sands

Modulus subgrade reaction (ks) kPa/m

Table 18. Modulus Subgrade Reaction and Soil Strain Value for Clay (after Reese, 1987)Relative DensityLooseMediumDense

Submerged Sand54301630033900

Sand Above Water Table67902443061000

In general, the coefficient of subgrade reaction which is also known as the modulus of subgrade reaction, or the subgrade modulus could be approached by using formula below :

Where:

ks= coefficient of subgrade reaction (kPa/m)

q= bearing pressure (kN/m2 or kPa)

(= settlement (m)Basic TheoryShallow Foundation

Shallow foundations transmit the applied structural loads to the near-surface soils. In the process of doing so, they induce both compressive and shear stresses in these soils. The magnitudes of these stresses depend largely on the bearing pressure and the size of the footing. If the bearing pressure is large enough, or the footing is small enough, the shear stresses may exceed the shear strength of the soil or rock, resulting in a bearing capacity failure.

Researchers have identified three types of bearing capacity failures:

General shear failure

It occurs in soils that are relatively incompressible and reasonably strong, in rock, and in saturated, normally consolidated clays that are loaded rapidly enough that the undrained condition prevails. The failure surface is well defined and failure occurs quite suddenly. A clearly formed bulge appears on the ground surface adjacent to the foundation. Although bulges may appear on both sides of the foundation, ultimate failure occurs on one side only, and it is often accompanied by rotations of the foundation. Local shear failure

Local shear failure is an intermediate case. The shear surfaces are well defied under the foundation, and then become vague near the ground surface. A small bulge may occur, but considerable settlement, perhaps on the order of half the foundation width, is necessary before a clear shear surface form near ground. Even then, a sudden failure does not occur, as happens in the general shear case. The foundation just continues to sink ever deeper into the ground.

Punching shear failure

The opposite extreme is the punching shear failure. It occurs in very loose sands, in a thin crust of strong soil underlain by a very weak soil, or in weak clays loaded under slow, drained conditions. The high compressibility of such soil profiles causes large settlements and poorly defined vertical shear surfaces. Little or no bulging occurs at the ground surface and failure develops gradually.

Figure 47. (a) General shear failure; (b) Local shear failure; (c) Punching shear failure (after Vesic, 1973)Complete quantitative criteria have yet to be developed to determine which of these three modes of failure will govern in any given circumstance, but the following guidelines are helpful: Shallow foundations in rock and undrained clays are governed by the general shear case

Shallow foundations in dense sands are governed by the general shear case. In this context, a dense sand is one with a relative density, Dr, greater than about 67%

Shallow foundations on loose to medium dense sands (30% < Dr < 67%) are probably governed by local shear

Shallow foundations on very loose sand (Dr < 30%) are probably governed by punching shear.

For nearly all practical shallow foundation design problems, it is only necessary to check the general shear case, and then conduct settlement analyses to verify that the foundation will not settle excessively. These settlement analyses implicitly protect against local and punching shear failures.Bearing Capacity Analyses in Soil General Shear CaseUltimate capacity for shallow foundation design is calculated by using the formula taken from Terzaghi (1943):

(for continuous foundation)

(for square foundation)

(for circular foundation)Where:

c= effective cohesion of soil

(= effective unit weight of soil

q = vertical effective overburden pressure

Nc, Nq, N( = bearing capacity factorsThe Terzaghi bearing capacity factors are:

for ( = 0

for ( > 0

The formula developed in Vesic (1973, 1975) is based on theoretical and experimental findings from these and other sources and is an excellent alternative to Terzaghi. It produces more accurate bearing values and it applies to a much broader range of loading and geometry conditions. The primary disadvantage is its added complexity.

Vesic retained Terzaghis basic format and added the following additional factors:

sc, sq, s(= shape factors

dc, dq, d(= depth factors

ic, iq, i(= load inclination factors

bc, bq, b(= base inclination factors

gc, gb, g(= ground inclination factors

He incorporated these factors into the bearing capacity formula as follows:

Terzaghis formulas consider only vertical loads acting on a footing with a horizontal base with a level ground surface, whereas Vesic factors allow any or all of these to vary. The notation for these factors is shown in figure below:

Figure 48. Notation for Vesics load inclination, base inclination, and ground inclination factors. All angles are expressed in degreesShape FactorsVesic considered a broader range of footing shapes and defined them in his s factors:

For continous footings, B/L ( 0, so sc, sq, and s( become equal to 1. This means the s factors may be ignored when analyzing continous footings.

Depth Factors

Unlike Terzaghi, Vesic has no limitations on the depth of the footing. The depth of footing is considered in the following depth factors:

For relatively shallow foundations (D/B 1), use . For deeper footings (D/B > 1), use with the tan-1 term expresses in radians. Note that this produces a discontinues function at .Load Inclination Factors

The load inclination factors are for loads that do not act perpendicular to the base of the footing, but still act through its centroid. The variable P refers to the component of the load that acts perpendicular to the bottom of the footing, and V refers to the component that acts parallel to the bottom.

The load inclination factors are:

For loads inclined in the B direction:

For loads inclined in the L direction:

Where:

V= applied shear load

P= applied normal load

A= base area of footing

cu= cohesion

(= friction angle

B= foundation width

L= foundation length

If the load acts perpendicular to the base of the footing, the i factors equal 1 and may be neglected. The i factors also equal 1 when ( = 0.Base Inclination FactorsThe vast majority of footings are built with horizontal bases. However, if the applied load is inclined at a large angle from the vertical, it may be better to incline the base of the footing to the same angle so the applied load acts perpendicular to the base. However, keep in mind that such footings may be difficult to construct.

The base inclination factors are:

If the base of the footing is level, which is the usual case, all of the b factors become equal to 1 and may be ignored.Ground Inclination Factors

Footings located near the top of a slope have a lower bearing capacity than those on level ground. Vesic ground inclination factors, presented below, account for this. However, there are also other considerations when placing footings on or near slopes.

If the ground surface is level (( = 0), the g factors become equal to 1 and may be ignored.Bearing Capacity FactorsVesic used the following formulas for computing the bearing capacity factors Nq, Nc and N( :

for ( > 0

for ( = 0

Bearing Capacity Analysis in Soil Local and Punching Case

Engineers rarely need to compute the local or punching shear bearing capacities because settlement analyses implicity protect against this type of failure. In addition, a complete bearing capacity analysis would be more complex because of the following:

These modes of failure do not have well-defined shear surfaces, such as those shown in figures above, and are therefore more difficult to evaluate.

The soil can no longer be considered incompressible (Ismael and Vesic, 1981).

The failure is not catastrophic, so the failure load is more difficult to define.

Scale effects make it difficult to properly interpret model footing tests.

Terzaghi (1943) suggested a simplified way to compute the local shear bearing capacity using the general shear formulas with appropriately reduced values of c and (:

Vesic (1975) expanded upon this concept and developed the following adjustment formula for sands with a relative density, Dr, less than 67%:

Where:cadj= adjusted cohession(adj= adjusted friction anglec= cohession

(= friction angleDr= relative density of sand, expressed in decimal form (0 Dr 67%)

Although the Vesic (1975) formula was confirmed with a few model footing tests, both methods are flawed because the failure mode is not being modeled correctly. However, local or punching shear will normally only govern the final design with shallow, narrow footings on loose sands, so an approximate analysis is acceptable. An important exception to this conclusion is the case of a footing supported by a thin crust of strong soil underlain by very weak soil. This would likely be governed by punching shear and would justify a custom analysis.

Sliding Capacity AnalysisResistance against sliding in soils :

Where :

Fs = minimum factor of safety against sliding potential

W = foundation weight including soil above footing

PV = vertical force

PH = horizontal force

Pp = passive force

Abase = base areaca = adhesion factortan = friction factor between soil and base

(= friction angle

tan (= tan Table 19. Adhesion factor (NAVFAC DM-7.02)Interface Materials (Cohesion)Adhesion, ca (kPa)

Very soft cohesive soil (0 12.5 kPa)0 12.5

Soft cohesive soil (12.5 20 kPa)12.5 20

Medium stiff cohesive soil (20 50 kPa)20 37.5

Stiff cohesive soil (50 100 kPa)37.5 47.5

Very stiff cohesive soil (100 200 kPa)47.5 65

Pile FoundationBearing capacity calculation can be determined by using analysis of such this condition:

1. Bearing capacity of single pile

2. Pile-soil interaction and pile group

Figure 49. Pile Foundation Analysis

Bearing Capacity of Single Pile

Figure 50. Bearing Capacity of PileIn general, the ultimate axial capacity of bored pile / driven pile can be obtained as the summation of end bearing capacity plus skin friction resistance, or :

Where:

Qu = ultimate pile capacity

Qp = ultimate end bearing capacity

Qs = ultimate skin friction resistenceAnalyses Based on SPT Results

Toe-BearingThe ultimate end bearing capacity (Qp) can be determined as:

Where:

Qp= ultimate end bearing capacity

qp = unit end bearing capacity

Ap= area of bored pile / driven pile cross section

The calculation of unit end bearing capacity (qp) must refer to the soil condition. Below are several equations for its conditions:

Because of their low hydraulic conductivity, we assumed undrained conditions exist in clays beneath the toe of deep foundation. Therefore, we compute qp using the undrained shear strength, cu. For deep foundations which have ratio D/B > 3 with cu 250 kPa :

Where:

qp = unit end bearing capacity

Nc* = bearing capacity factor (ONeill and Reese, 1999)

= 6,5 at cu = 25 kPa

= 8,0 at cu = 50 kPa

= 9,0 at cu ( 100 kPa

cu= undrained shear strength in the soil between the base of the shaft/pile and a distance 2Bb below the base

D = depth to the bottom of the shaft/pileBb = diameter of shaft/pile base Clays with cu > 250 kPa should be evaluated as intermediate geo-materials.

If the base diameter, Bb, is greater than 1900 mm, the value of qp could produce settlements greater than 25 mm, which would be unacceptable for most buildings. To keep settlements within tolerable limits, reduce the value of qp and use this value (ONeill and Reese, 1999):

Wher