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REVIEW OF METHODS TO ENHANCE THE DUCTILITY AND STRENGTH OF STRUCTURES FOR RESISTING EARTHQUAKE

AND BLAST LOADING

Ravi Sudhakar Vemulapalli

Problem report submitted to the College of Engineering and Mineral Resources

at West Virginia University in partial fulfillment of the requirements

for the degree of

Master of Science in

Civil Engineering

Udaya B. Halabe, Ph.D., P.E., Chair Hema J. Siriwardane, Ph.D., P.E.

Roger H. L. Chen, Ph.D.

Department of Civil and Environmental Engineering

Morgantown, West Virginia 2007

Keywords: Earthquake, Blast, Blast Design, Seismic Design, Seismic Rehabilitation,

Blast Rehabilitation, FRP, Wrapping, Jacketing

ABSTRACT

Review of Methods to Enhance the Ductility and Strength of Structures for Resisting Earthquake and Blast Loading

Ravi Sudhakar Vemulapalli

This report presents a literature review of the methods to enhance the ductility and

strength of structures to resist earthquake and blast loadings. Enhancing the ductility of

the structural system greatly helps in minimizing the damage to the structure in the event

of earthquake or blast loading. This literature review consists of methods like wrapping

beam column joints with FRP composites, use of stay-in-place FRP confinements for RC

columns, use of shear walls, RC jacketing, base isolation and energy dissipation devices,

and retrofitting of unreinforced masonry walls. Most of the existing buildings were

designed based on earlier design codes that are no longer in practice and don’t account

for the required ductility that is needed for efficient seismic and blast resistance. The

ductility and strength of the structures can be improved through effective rehabilitation

techniques, which in turn help in effective mitigation of damage due to earthquake and

blast loading. This report also includes a review of techniques especially aimed at

retrofitting commercial buildings to resist blast loading.

All the methods that were reviewed are effective in enhancing structural resistance to

seismic and blast loading, but some of them have applicability limitations. More research

work is needed to further develop these techniques and make them more cost effective.

ACKNOWLEDGEMENTS

At the outset, I would like to express my sincere gratitude to my academic advisor, Dr.

Udaya B. Halabe for his valuable guidance and encouragement during my Master of

Science in Civil Engineering (M.S.C.E) degree program. I am thankful to Dr. Hema J.

Siriwardane and Dr. Roger Chen, members of the Advisory and Examining Committee,

for their help during my studies.

I am thankful to Shasanka Dutta, Aneesh Bethi, and Sandeep Pyakurel for their support

and encouragement during my work. I thank my family who stood beside me and

provided the impetus to do my best.

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TABLE OF CONTENTS

ABSTRACT ii ACKNOWLEDGEMENTS iii TABLE OF CONTENTS iv LIST OF FIGURES vii LIST OF TABLES x Chapter 1 – INTRODUCTION 1

1.1. BACKGROUND 1 1.2. OBJECTIVE 3 1.3. REPORT ORGANIZATION 3

Chapter 2 - EARTHQUAKE AND BLAST 4 2.1. EARTHQUAKE 4 2.1.1. Types of Earthquake 4 2.1.2. Quantifying Earthquake 4 2.1.3. Different Types of Structural Failures 6 2.1.4. Important Categories of Damage 9 2.1.4.1. Reinforced Concrete Members 10 2.1.4.2. Structural Steel Members 13 2.1.4.3. Masonry Structures 13 2.1.5. Previous Earthquakes 14 2.2. BLAST 16 2.2.1. Blast Resistant Design 16 2.2.2. Types of Explosions 17 2.2.3. Blast Wave Parameters 20 2.2.4. Previous Explosions 21 Chapter 3 - REHABILITAION FOR EARTHQUAKE 24 3.1. BEAM-COLUMN JOINT REHABILITATION 24 3.1.1. Experimental Program 25 3.1.1.1. Specimen Description 25 3.1.1.2. Material Properties 26 3.1.1.3. Experimental Test Set-Up 27 3.1.1.4. Loading Sequence 27 3.1.1.5. Instrumentation 29 3.1.2. Rehabilitation Scheme 31 3.1.3. Experimental Results 32 3.1.4. Summary 33 3.2. SQUARE HIGH-STRENGTH CONCRETE COLUMNS IN FRP

STAY-IN-PLACE FORMWORK 34

3.2.1. Experimental Procedure 34 3.2.2. Test Specimens 35 3.2.3. Material Properties 39 3.2.3.1. Carbon FRP composite 39

iv

3.2.3.2. Concrete Properties 41 3.2.3.3. Steel Reinforcement 41 3.2.4. Experimental Test Setup 42 3.2.5. Test Results 43 3.2.6. Deformation of HSC Columns Confined by FRP

Casing 44

3.2.7. Summary 45 3.3. REHABILITATION OF COLUMNS WITH REINFORCED

CONCRETE JACKETS 45

3.3.1. Added Longitudinal Reinforcement 45 3.3.2. Slab Crossing 47 3.3.3. Interface Surface Preparation 47 3.3.3.1. Methods to Increasing Surface Roughness 47 3.3.3.2. Surface Pre-wetting 48 3.3.3.3. Application of Bonding Agents 49 3.3.3.4. Addition of Steel Connectors 50 3.3.3.5. Testing of Different Methods 50 3.3.4. Spacing of Added Stirrups 51 3.3.5. Temporary Shoring of the Structure 51 3.3.6. Properties of Added Concrete 51 3.3.7. Structural Behavior 52 3.3.7.1. Effect of Damage on Structural Behavior 53 3.3.8. Summary 53 3.4. SEISMIC REHABILITATION BY ADDING SHEAR WALLS 55 3.4.1. Rehabilitation Method 55 3.4.2. Design criteria for seismic design 55 3.4.3. Four Story Building in Dinar 56 3.4.4. Eight Story Building in Ceyhan 58 3.4.5. Aftershock Test in Dinar 60 3.5. PASSIVE SEISMIC PROTECTION IN STRUCTURAL

REHABILITATION 61

3.5.1. Base Isolation 62 3.5.2. Energy Dissipation Devices 64 Chapter 4 - REHABILITAION FOR BLAST 66 4.1. POLYMER RETROFIT OF UNREINFORCED MASONRY WALLS 67 4.1.1. Selection of Retrofit Material 69 4.1.2. Test Procedures 70 4.1.3. Instrumentation 71 4.1.4. Test Results 76 4.1.5. Summary 76 4.2. REHABILITATION OF STRUCTURE AFTER A GAS EXPLOSION 76 4.2.1. Rehabilitation Method 77 4.2.2. Case Study of a Structure Damaged by Gas Explosion 78 4.2.3. Building Damage Assessment 79 4.2.4. Rehabilitation Scheme 81

v

4.2.5. Summary 86 4.3. BLAST RESISTANT DESIGN OF COMMERCIAL BUILDINGS 86 4.3.1. External Treatment 89 4.3.1.1. Stand off Distance 89 4.3.1.2. Lower Floor Exterior 90 4.3.2. Glazing 91 4.3.3. Facade and Atrium 92 4.3.3.1. Exterior of the Atrium 92 4.3.3.2. Interior of Atrium 92 4.3.4. Floor Slabs 93 4.3.5. Columns 96 4.3.6. Transfer Girders 99 Chapter 5 – CONCLUSIONS AND RECOMMENDATIONS 100 5.1. CONCLUSIONS 100 5.2. RECOMMENDATIONS 103 References 104

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LIST OF FIGURES

Figure 1.1.1: Trends of the global disasters (Adramovitz 2001) 1 Figure 1.1.2: Distribution of global deaths by disasters 1985-99 (Adramovitz 2001) 2 Figure 2.1.1: Schematic diagram of Earthquake (Booth et al. 2006) 5 Figure 2.1.2: Multi story reinforced concrete structure collapsed in Mexico City,

1985 (Booth et al. 2006) 7

Figure 2.1.3: Ground story collapse (soft story) of a building during the Turkey earthquake in Erzincan, Turkey (Booth et al. 2006)

7

Figure 2.1.4: Beginning of a soft-story collapse of a building in Erzincan during the 1992 Turkey earthquake (Booth et al. 2006)

8

Figure 2.1.5: Upper story collapse of a multi story structure during the 1985 Mexico City earthquake (Booth et al. 2006)

8

Figure 2.1.6: Intermediate story failure in Hotel Decare Building during 1985 Mexico City earthquake (Booth et al. 2006)

9

Figure 2.1.7: Parapet wall failure on a building during 2001 Gujarat, India earthquake (Booth et al. 2006)

10

Figure 2.1.8: Beam Column Joint Failure in Erzincan, Turkey during the 1992 earthquake (Booth et al. 2006)

11

Figure 2.1.9: Bursting failure of a column in Northridge, California during 1994 earthquake (Booth et al. 2006)

11

Figure 2.1.10: Shear failure of a column in St. Johns, Antigua during 1974 earthquake (Booth et al. 2006)

12

Figure 2.1.11: Out of plane failure of an unreinforced masonry wall (Booth et al. 2006)

13

Figure 2.1.12: View of the collapsed Margalla Towers after the 2005 earthquake (Pakistan earthquake web site 2007)

14

Figure 2.1.13: Damage caused by the 1994 Northridge earthquake (Northridge earthquake web site 2007)

15

Figure 2.1.14: Damage caused by the 1994 Northridge earthquake (Northridge earthquake web site 2007)

16

Figure 2.2.1: Mechanical explosion in an Industry (Explosions web site 2007) 18 Figure 2.2.2: Structure affected by sewer explosion (Explosions web site 2007) 18 Figure 2.2.3: Building affected by LP gas explosion (Explosions web site 2007) 19 Figure 2.2.4: Cloud formed after nuclear explosion (Explosions web site 2007) 19 Figure 2.2.5: Shock wave and pressure wave (ASCE 1997) 20 Figure 2.2.6: Damaged stories of the Alfred P. Murrah Federal Building, Oklahoma.

1995 (Oklahoma City web site 2007) 22

Figure 2.2.7: Damages to building and the surroundings after the 1995 Oklahoma City bombings (Oklahoma web site 2007)

22

Figure 2.2.8: WTC tower one after the 1993 explosion in the parking lot (WTC web site 2007)

23

Figure 3.1.1: Reinforcement details of the beam-column specimen (Ghobarah et al. 2001)

26

Figure 3.1.2: Experimental set up for the beam-column joint testing (Ghobarah et al. 2001)

28

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Figure 3.1.3: Cyclic loading applied to the free end of the beam (Ghobarah et al. 2001)

29

Figure 3.1.4: Location of strain gauges on the reinforcement. (Ghobarah et al. 2001) 30 Figure 3.1.5: Rehabilitation scheme for the beam-column joint (Ghobarah et al.

2001) 30

Figure 3.1.6: Beam-column joint with the FRP rehabilitation (Ghobarah et al. 2001) 31 Figure 3.1.7: FRP laminate failure of specimen T1R (Ghobarah et al. 2001) 33 Figure 3.2.1: Geometry of the columns specimen used in the testing (Ozbakkalogu

et al. 2007) 36

Figure 3.2.2: Reinforcement arrangement used in specimen RS-1 (Ozbakkalogu et al. 2007)

37


37


38


38


39


39

Figure 3.2.8: Wooden templates used to make the FRP casings (Ozbakkalogu et al. 2007)

40

Figure 3.3.1: Failure of the steel bars of the column and slippage of the steel bars of the added jacketing (Julio et al. 2003)

46

Figure 3.3.2: Specimens prepared by sand-blasting (Julio et al. 2003) 48 Figure 3.3.3: Application of epoxy resin on specimen (Julio et al. 2003) 49 Figure 3.3.4: Steel connectors epoxy bonded on push off specimens (Julio et al.

2003) 50

Figure 3.3.5: Concrete casting of the jacket (Julio et al. 2003) 52 Figure 3.4.1: A four story building that’s been rehabilitated in Dinar (Sucuoglu et

al. 2004) 56

Figure 3.4.2: The Rehabilitated scheme of the four stories structure in Dinar, ground floor plan (Sucuoglu et al. 2004)

58

Figure 3.4.3: Eight story damaged building in Ceyhan (Sucuoglu et al. 2004) 59 Figure 3.4.4: Rehabilitation scheme of the eight stories building in Ceyhan, ground

floor plan (Sucuoglu et al. 2004) 60

Figure 3.5.1: Different types of base isolation systems (Guerreiro et al. 2006) 63 Figure 3.5.2: Example of retro-fitting a RC building with base isolation (Guerreiro

et al. 2006) 63

Figure 3.5.3: Los Angeles City Hall rehabilitated with base isolators (Guerreiro et al. 2006)

64

Figure 3.5.4: Arrangement of a viscoelastic damper (Guerreiro et al. 2006) 65 Figure 4.1.1: Test setup for testing the effectiness of the spary-on polymer

retrofitting method (Davidson et al. 2004) 69

Figure 4.1.2: Instrumentation plan (Davidson et al. 2004) 70

viii

Figure 4.1.3: Test 1 reflected pressure: gauge R1 and R2 (Davidson et al. 2004) 71 Figure 4.1.4: View of the damaged walls after test 1 (Davidson et al. 2004) 72 Figure 4.1.5: Test 3 Result: wall panels on reaction structure (Davidson et al. 2004) 74 Figure 4.1.6: Test 3 Result: Wall panels on test cubicles (Davidson et al. 2004) 75 Figure 4.1.7: Tearing of the polymer on the inside from Test 3 (Davidson et al.

2004) 76

Figure 4.2.1: Typical floor plan of the structure (Bob 2004) 79 Figure 4.2.2: View of the damaged transverse walls (Bob 2004) 80 Figure 4.2.3: View of the damaged floor-wall connection (Bob 2004) 80 Figure 4.2.4: View of the damaged RC floor (Bob 2004) 81 Figure 4.2.5: strengthening with RC Coating (Bob 2004) 82 Figure 4.2.6: strengthening of walls with CFRP (Bob 2004) 83 Figure 4.2.7: Installation of CFRP to strengthen floors (Bob 2004) 83 Figure 4.2.8: Strengthening of floors with CFRP (Bob 2004) 84 Figure 4.2.9: Strengthening with RC sheets, columns and longitudinal beams (Bob

2004) 85

Figure 4.2.10: Local strengthening of columns and beams with RC jacketing (Bob 2004)

85

Figure 4.2.11: Strengthening of floors with CFRP (sika wrap) (Bob 2004) 86 Figure 4.3.1: Elevation view of the building (Ettouney et al. 1996) 87 Figure 4.3.2: Plan view of the building (Ettouney et al. 1996) 88 Figure 4.3.3: Isometric view of the building (Ettouney et al. 1996) 88 Figure 4.3.4: Failure mechanism of the flat slab (Ettouney et al. 1996) 93 Figure 4.3.5: Effects of blast loading on columns (Ettouney et al. 1996) 94 Figure 4.3.6: Lateral load carrying mechanism and effects after the blast (Ettouney

et al. 1996) 94

Figure 4.3.7: High and low vulnerability locations (Ettouney et al. 1996) 95 Figure 4.3.8: Flat slab improvements (Ettouney et al. 1996) 95 Figure 4.3.9: Direct lateral loading of column (Ettouney et al. 1996) 97 Figure 4.3.10: Uplifting on the columns (Ettouney et al. 1996) 97 Figure 4.3.11: Progressive collapse mechanism (Ettouney et al. 1996) 99

ix

LIST OF TABLES Table 3.1.1: GFRP material properties. (Ghobarah et al. 2001) 27 Table 3.2.1: FRP casing and crosstie details (Ozbakkalogu et al. 2007) 42 Table 3.2.1: Reinforcement details (Ozbakkalogu et al. 2007) 42 Table 4.1.1: Average tensile strength properties obtained from the tests (Davidson

et al. 2004) 67

Table 4.1.2: Properties of Polyurea (Davidson et al. 2004) 68 Table 4.1.3: Material Properties obtained for the tests (Davidson et al. 2004) 68 Table 4.1.4: Gauge measurements obtained from Test 1 (Davidson et al. 2004) 71 Table 4.1.5: Gauge measurements obtained from Test 2 (Davidson et al. 2004) 73 Table 4.1.6: Gauge measurements obtained from Test 3 (Davidson et al. 2004) 75 Table 5.1.1: Rehabilitation techniques for resisting earthquake loading 101Table 5.1.2: Rehabilitation techniques for resisting blast loading 102

x

Chapter 1

INTRODUCTION

1.1 BACKGROUND

Over the years the total number of disasters is continually increasing. During the last

century the number of deaths due to disasters, which include earthquake, flood, wind

storm, hurricane, tornado, etc., was more than 10 million. The economic losses due to the

catastrophes has also been rising, as is evident from Figure 1.1.1 where it is seen that the

economic losses rose from nearly 40 billion dollars in 1950’s to 600 billion dollars in

1990’s. Of all the disasters, earthquake accounts for nearly one-third of the total global

deaths (Figure 1.1.2).

Figure 1.1.1: Trends of the global disasters (Adramovitz 2001)

Earthquake is a very common phenomenon in many parts of the United States. The major

areas of significant seismic hazard in the US are California, the Pacific North West,

Nevada, Idaho, Montana, Utah and Colorado. There is also a significant risk in the

Central, Northeastern and Southeastern part of the country. Over the years, US has

witnessed myriads of major earthquakes that resulted in significant property loss and

1

death toll. Examples include the 1811-12 New Madrid earthquake (Richter magnitude of

8.0 – 8.6), the 1971 San Fernando Valley earthquake (Richter magnitude of 6.6), the

1979 El Centro earthquake which is also known as the Imperial Valley earthquake

(Richter magnitude of 6.6), the 1989 San Francisco California earthquake (Richter

magnitude of 7.8), and 1994 Northridge earthquake (Richter magnitude of 6.7). The 1994

Northridge earthquake resulted in a property loss of over 20 billion and over 112,000

older building were damaged. The 1989 San Francisco earthquake caused a property

damage of 524 million dollars and the death toll of 3000. The 1886 Charleston, South

Carolina earthquake killed about 60 people and the property damage was estimated to be

6 million in 1999 dollars (Adramovitz 2001).

Figure 1.1.2: Distribution of global deaths by disasters 1985-99 (Adramovitz 2001)

The design of the existing buildings that were built as per the old design codes did not

take into account the issues of ductility. Even the latest codes cannot guarantee a

complete protection against earthquakes but they do minimize the effects. Most of the

older buildings are at risk in the event of an earthquake. Since it is not economically

feasible to reconstruct all the old buildings as per the new seismic design codes, the only

feasible solution would be rehabilitating these structures. This will improve the ductility

characteristics of the structure for resisting earthquakes.

2

Over the years the number of explosions in commercial buildings and public facilities is

also increasing at a rapid rate. Examples include the Oklahoma City blast in 1995, the

Word Trade Center blast in 1993, Bombay (India) blast of 1993 and The Twin Tower

collapse of 2001. Although blast loading has not been given much importance in civil

engineering design, due to the frequency of recent occurrences it is advisable to consider

blast loading in the design of structures thereby improving their response to blast

loadings.

Many rehabilitation techniques for enhancing the ductility and strength of structures to

resist earthquake and blast loading have been developing over the years. Some of these

techniques include wrapping beam column joints with Fiber Reinforced Polymer (FRP)

composites, use of stay-in-place FRP confinements for RC columns, use of shear walls,

RC jacketing, base isolation and energy dissipation devices, and retrofitting of

unreinforced masonry walls. In this problem report an attempt has been made to review

these methods.

1.2 OBJECTIVE

The main objective of this report is to conduct a literature review on various

rehabilitation techniques used for enhancing the structural resistance to earthquake and

blast loading. The review discusses rehabilitation techniques such as wrapping beam

column joints with FRP composites, use of stay-in-place FRP confinements for RC

columns, use of shear walls, RC jacketing, base isolation and energy dissipation devices,

and retrofitting of unreinforced masonry walls.

1.3 REPORT ORGANIZATION

This problem report consists of five chapters. The first chapter presents a brief

background, objectives and scoop of the study. Chapter 2 briefly discusses earthquakes,

earthquake damages, blast, blast damages and past experiences. Chapter 3 presents an

extensive literature review of earthquake rehabilitation techniques. Blast rehabilitation

techniques are reviewed in Chapter 4. Chapter 5 summarizes the conclusion and

recommendations. This is followed by a list of references cited in the text.

3

Chapter 2

EARTHQUAKE AND BLAST

This chapter provides a brief introduction to earthquake and blast loadings and also

discusses various details like the types of earthquakes, failure during earthquakes,

examples of past earthquakes, types of explosions, pressure waves, blast wave

parameters, and examples of previous explosions.

2.1 EARTHQUAKE

Earthquakes occur due to the forces within the earth crest to displace mass of rock

relative to each other. Stress is imposed on the lithosphere when the earth plates move.

When the stress is large and exceeds the capacity, lithosphere breaks or shifts. When

these plates move, they generate forces and when the forces are large they force the crust

to break. When the crust is cracked, stress is released in the form of energy which moves

along the surface of the earth as a wave, which we feel as earthquake (Booth et al. 2006).

2.1.1 Types of Earthquake

There are different types of earthquakes: tectonic, volcanic, collapse and explosion. The

type of earthquake that occurs depends on the geographical location and the make-up of

the region. The most common type of earthquake is the tectonic earthquake. Tectonic

earthquakes are caused due to the cracking of the earth’s crust by forces created by the

movement of the tectonic plates. Volcanic Earthquakes occur in conjunction with

volcanic activity. Collapse earthquake are relatively small earthquakes and occur at

locations near to mines and cravens. The major effect of these kinds of earthquakes is the

collapse of the roof of the craven or mine. Explosion earthquake occurs when a nuclear or

chemical bomb is detonated in a bore hole underground (Booth et al. 2006).

2.1.2 Quantifying Earthquake

There are two fundamental measures of an earthquake: earthquake magnitude and

earthquake intensity. Earthquake magnitude is the fundamental property of the

4

earthquake; it is the amount of energy released as measured on a logarithmic scale

(Richter Magnitude Scale). Earthquake intensity, as measured by the Modified Mercalli

(MM) Intensity Scale, on the other hand depends on the location of measurement. It

describes the effect of the earthquake on the people and building. In other words, the MM

scale is a measure of damage caused by an earthquake. As the distance from the epicenter

increases, the intensity of the earthquake decreases. As shown in the Figure 2.1.1 an

earthquake with a specific magnitude can have different intensities at different location as

the epicentral distance changes.

The two most common scales to measure the magnitude of the earthquake are body wave

magnitude mb (suitable for small magnitude events) and surface wave magnitude Ms

(most suitable for large events). Both measurements are measured by seismographs,

which measure ground tremors from a great distance. The third scale is the moment

magnitude, Mw, this is directly related to the amount of energy released in an earthquake.

This is suitable for all kinds of earthquakes (Booth et al. 2006).

Figure 2.1.1: Schematic diagram of Earthquake (Booth et al. 2006)

The displacement of the ground is denoted by ug and the displacement of the mass

relative to the ground is denoted as u. Then the total displacement of the mass is denoted

as ut (Chopra 2005).

5

ut(t) = u(t) + ug(t)

Both displacements ug and ut refer to the same inertial frame of reference and their

positive directions coincide. The relative motion u between the mass and the base due to

the structural deformation generates elastic and damping forces. When we consider the

concept of dynamic equilibrium between inertial, damping, and elastic forces:

fI + fD + fS = 0

The inertial force on the system is related to the mass and the acceleration on the mass ü

(Chopra 2005).

fI = m ü

Thus, the equation of dynamic equilibrium changes as follows

mü + ců + ku = - müg

This equation shows that the system behaves in the same way as when an external force

of - müg(t) is applied or a ground acceleration of - üg(t) is applied. Thus the relative

displacement u(t) of the structure due a ground acceleration of -üg(t) is similar to the

relative displacement u(t) due to an external force - müg (Chopra 2005).

2.1.3. Different Types of Structural Failures

Figures 2.1.1 to 2.1.5 show some of the types of structural failure in different parts of the

world during the last decade.

6

Figure 2.1.2: Multi story reinforced concrete structure collapsed in Mexico City, 1985

(Booth et al. 2006)

Figure 2.1.3: Ground story collapse (soft story) of a building during the Turkey

earthquake in Erzincan, Turkey (Booth et al. 2006)

7

Figure 2.1.4: Beginning of a soft-story collapse of a building in Erzincan during the 1992

Turkey earthquake (Booth et al. 2006)

Figure 2.1.5: Upper story collapse of a multi story structure during the 1985 Mexico City

earthquake (Booth et al. 2006)

8

2.1.4 Important Categories of Damage

When two buildings are very close to each other, during an earthquake two adjacent sides

of the buildings may pound into each other. This may cause a major damage, particularly

when the floor levels of the adjacent buildings are different. This is shown in Figure 2.1.6

where an intermediate story of the Hotel Decare was completely damaged. Appendages

to the building such as masonry parapets, cantilevers, roof tanks and pent houses behaved

badly during an earthquake. There are two reasons for this behavior. First reason is that

these structural members were not designed with adequate ductility. Second reason is the

effect of the dynamic amplification by the building to which they are attached (Figure

2.1.7).

Figure 2.1.6: Intermediate story failure in Hotel Decare Building during 1985 Mexico

City earthquake (Booth et al. 2006)

9

Figure 2.1.7: Parapet wall failure on a building during 2001 Gujarat, India earthquake

(Booth et al. 2006)

2.1.4.1 Reinforced Concrete Members

Buildings that consists of reinforced concrete beams and columns and which are not

properly braced by lateral walls providing lateral stiffness are very vulnerable to

earthquakes, unless special measures are taken to enhance the ductility of the ductility of

the structures (Booth et al. 2006).

The main vulnerabilities in RC structures are:

Failure of Beam-Column Joints (Figure 2.1.8)

Bursting Failure in Columns (Figure 2.1.9)

Shear Failure in Columns (Figure 2.1.10)

Soft Story Failure (Figure 2.1.3)

10

Figure 2.1.8: Beam Column Joint Failure in Erzincan, Turkey during the 1992


Figure 2.1.9: Bursting failure of a column in Northridge, California during 1994


11

Figure 2.1.10: Shear failure of a column in St. Johns, Antigua during 1974 earthquake

(Booth et al. 2006)

When inadequate transverse steel reinforcement is provided in a column, during an

earthquake the longitudinal bars carrying large axial loads and lateral loads buckle due to

insufficient confinement provided by the transverse reinforcement (Figure 2.1.9). In

Figure 2.1.10 the masonry wall stops below the full height of the column thereby creating

a short column liable to fail in shear rather than bending.

When buildings are built with their lower floor serving as the parking space without any

walls, the lateral stiffness of the building at the lower level is very low compared to the

upper levels. This causes the failure of the lower floor and ultimately a complete collapse

of the building. This is normally referred as a “Soft Story Failure”. Buildings with shear

walls have proven to be much more effective in resisting earthquakes compared to the

buildings without shear walls. The lateral stiffness of the buildings increase due to shear

walls and the buildings can resist earthquake loading better (Booth et al. 2006).

Ductility and Strength of structures and structural members play an important role in

withstanding earthquake forces. All frame elements must be designed in such a way that

12

they respond to a strong earthquake in a ductile fashion. Non-ductile modes such as shear

and bond failure must be avoided.

2.1.4.2 Structural Steel Members

Structural steel members exhibit the following types of failures during an earthquake.

Brittle failure of Bolts in shear or tension.

Member buckling.

Local web and flange buckling.

Large deflection of unbraced members.

Failure of steel members and other building elements (Booth et al. 2006).

2.1.4.3 Masonry Structures

Failure of both unreinforced and reinforced masonry is common. In-plane masonry is

very stiff so the forces transmitted by the ground motion are very high and masonry walls

are also very brittle so the failure in this case is a complete collapse or diagonal cracking

in both directions in the form of an “X”. Out of plane, free standing masonry walls are

highly vulnerable to earthquake and are liable to toppling failure (Figure 2.1.11).

Masonry walls that are mechanically connected on the sides and top of the wall are less

likely to have a toppling failure. Masonry walls with continues reinforcement are more

effective in avoiding total collapse (Booth et al. 2006).

Figure 2.1.11: Out of plane failure of an unreinforced masonry wall (Booth et al. 2006)

13

2.1.5 Previous Earthquakes

The Kashmir Earthquake (also known as the Great Pakistan Earthquake) was a major

earthquake with 7.6 magnitude on the Richter scale. The epicenter of the earthquake was

in Pakistan controlled Kashmir and the earthquake occurred at 8:50:38 Pakistan standard

time on October 8, 2005. The official death toll in Pakistan was 73,276, and 1,400 in

Jammu & Kashmir and 14 in Afghanistan.

One of the buildings famously known as the Margalla Towers (Figure 2.1.12), a 10 story

building, had collapsed completely in Islamabad killing most of the occupants. Although

the Margalla Towers was completely destroyed, neighboring buildings Al-Mustafa and

Park Towers stood their ground. This clearly points to the structural deficiency of the

collapsed building to withstand earthquake loads. The total property loss in the Pakistan

earthquake was around $2.3 Billion (Kashmir earthquake web site 2007).

Figure 2.1.12: View of the collapsed Margalla Towers after the 2005 earthquake

(Pakistan earthquake web site 2007)

14

The Great Chilean Earthquake of May 22, 1960, was a major earthquake with a Richter

magnitude of 9.5. It occurred in the early afternoon 19:11 (UTC) and this resulted in a

tsunami that affected southern Chile, Argentina, Hawaii, Japan, the Philippines, eastern

New Zealand and the Aleutian Islands in Alaska. The epicenter of the earthquake was in

Valdivia, Chile 700 kilometers south of Santiago. This caused localized tsunami waves

that reached heights of 25m. These waves were 10.7 meter at a distance of 10,000

kilometers from the epicenter. The total fatalities in this earthquake as published in the

USGS citing studies state figures of 2231, 3000, or 5700 killed. The total estimated

monetary costs are 400 million to 800 million dollars (after adjusting for inflation, it

would be around 2.6 to 5.2 billion in 2005 dollars) (Chile earthquake web site 2007).

The Northridge earthquake occurred on January 17, 1994. This earthquake hit the city of

Los Angeles, California at 4:31 AM and the magnitude of the earthquake was 6.7 on the

Richter scale. The ground acceleration recorded is the maximum that’s ever recorded in

urban area in North America. This earthquake produced unusually strong ground

acceleration in the range of 1.0g. The total death toll in this earthquake was 72 people and

the estimated property damage was $12.5 billions. Damages were also caused by fire and

landslides. In terms of property damage this earthquake is one of the worst natural

disasters in the United States (Northridge earthquake web site 2007).

Figure 2.1.13: Damage caused by the 1994 Northridge earthquake (Northridge

earthquake web site 2007)

15

Figure 2.1.14: Damage caused by the 1994 Northridge earthquake (Northridge

earthquake web site 2007)

2.2 BLAST

The need and requirements for blast resistant design has evolved in the recent years, in

the wake of unexpected events, like the terrorist attacks. The main focus is on the

structural aspects of design and rehabilitation of buildings for blast loadings. The blast

resistant design of buildings is one of the measures to minimize the risk to people and

facilities from the hazards of explosions.

2.2.1 Blast Resistant Design

The primary objectives of blast resistant design are to achieve improved personnel safety

and minimize financial losses. The primary goal in blast resistant design is to provide the

personnel in the building with the same level of safety as that of people outside the

building. In the recent past we have seen that most of the casualties in an event of a blast

are due to the collapse of the buildings, collapse of structural members or the material

shattered by the blast wave. Therefore, the main aim is to reduce the probability that the

building itself becomes a hazard in an explosion. The other main concern is to prevent the

huge financial losses after a blast in terms of building reconstruction cost. Also, buildings

16

often contain important information or serve important function (e.g., hospitals, hotels).

Business information and “loss of use” often have a very high value, so destruction of

buildings could cause significant financial losses.

The requirements for the buildings are greatly influenced by factors of distance from the

source (stand-off distance) and expected occupancy. For example, a building that is

situated far enough from any potential bomb threat may not need increased blast

resistance. If maintaining high standoff distance is not possible, then a high level of blast

resistance should be provided to the building (ASCE 1997), but this increases the cost.

When a building does not have enough stand-off distance from a potential blast source,

the building is exposed to damaging overpressure, so blast resistant design is

recommended for theses structures to improve their resistance to blast loading.

2.2.2 Types of Explosions

There are four general types of explosions:

1) Mechanical explosion

2) Chemical explosion

3) Nuclear explosion

4) Electrical explosion

Mechanical explosions (Figure 2.2.1) are those in which vessel failure or rupture of the

container is created by high pressure gas. When the gas stored in the container are

flammable, then in many instances a resultant fire occurs as long as there is an ignition

source or the temperature is above the ignition temperature of the gas (Explosions web

site 2007).

17

Figure 2.2.1: Mechanical explosion in an Industry (Explosions web site 2007)

Chemical explosions occur by generation of high pressure gas as a result of exothermic

reaction resulting from initiation of chemical explosives or fuel cells. Some of the fuels

that may cause a chemical explosion are flammable gases, vapors of combustible gases,

carbon monoxide and carbon dioxide explosions.

Figure 2.2.2: Structure affected by sewer explosion (Explosions web site 2007)

18

Figure 2.2.3: Building affected by LP gas explosion (Explosions web site 2007)

Nuclear explosions occur from the high quantities of heat and gases released during a

fusion or fission processes.

Figure 2.2.4: Cloud formed after nuclear explosion (Explosions web site 2007)

High energy electric arcs may sometimes generate sufficient heat to start an explosion.

Some times the electric arcs can heat the surrounding gases to a high temperature and

cause a mechanical explosion. A typical example of this kind of explosion is an electric

panel box that has been violently dislodged from the rest of the box. This typically

19

happens during a lightening strike or a high energy arc. This reaction may or may not

cause subsequent fire (Explosions web site 2007).

2.2.3 Blast Wave Parameters

The most important feature of an explosion is the sudden release of energy to the

atmosphere and creating a pressure transient, or blast wave. The blast wave travels

outward in all direction from the source of the explosion. The magnitude and the shape of

the blast wave depend on nature of the energy released and on the distance from the

epicenter of the explosion (ASCE 1997).

There are two types of blast waves.

Shock Wave: This has a sudden rise in the pressure from the ambient pressure to a peak

free field overpressure. Then the pressure reduces gradually to the ambient pressure with

some highly damped pressure oscillations. This results in a negative pressure wave

following the positive blast wave (Figure 2.2.5a).

Pressure Wave: This has a gradual increase in the side-on pressures and reaches the peak

free field overpressure and then gradually reduces to the ambient pressure and a negative

phase (Figure 2.2.5b) in the same wave as the shock wave.

(a) Shock wave (b) Pressure wave

Figure 2.2.5: Shock wave and pressure wave (ASCE 1997)

20

The area under the pressure-time graph is called the Impulse of the wave. So, the Positive

impulse of the wave is calculated as

∫=td

O dttPI0

).(

= 0.5 PSO td, for a triangular wave

= 0.64 PSO td, for a half-sine wave

= c PSO td, for a exponentially decaying shock wave

where,

P(t) = overpressure function of time

PSO = Peak, side on, or incident, overpressure

td = duration of positive phase

c = a value between 0.2 and 0.5 depending on PSO.

When the blast wave encounters a surface it will reflect off of the surface and the surface

experiences a pressure greater than the peak overpressure. That value is called the

reflected pressure, which can be calculated by the following formula (ASCE 1997).

Pr = Cr PSO

where,

Cr = reflection coefficient (dependent on the peak over pressure, angle of incidence of the

wave front and the type of wave).

2.2.4 Previous Explosions

On April 19, 1995 the Alfred P. Murrah Federal Building in Oklahoma City was attacked

by terrorists. The explosion partially destroyed the structure (Figure 2.2.6). The attack

claimed 168 lives and almost 800 people were injured. Prior to the terrorist attack on

September 11, 2001 on the World Trade Center, this was the most deadly terrorist attack

in the US. Figures 2.2.6 and 2.2.7 show the structural damages incurred by the building.

More than the blast itself, the main cause of the casualties is the destruction of the front

part of the building. Since this building had an access road on the front of the building

there was not much stand off distance. Hence a proper blast resistant design of the rebuilt

structure will make it safer in case of an explosion (Oklahoma web site 2007).

21

Figure 2.2.6: Damaged stories of the Alfred P. Murrah Federal Building, Oklahoma.

1995 (Oklahoma City web site 2007)

Figure 2.2.7: Damages to building and the surroundings after the 1995 Oklahoma City

bombings (Oklahoma web site 2007)

22

On February 26, 1993 a truck bomb was planted in the underground parking garage in the

Tower One of the World Trade Center in New York. The original intentions of the

terrorists were to damage one side of the structure and in turn make Tower One to

collapse over the second tower destroying both the towers and causing lot of deaths and

huge financial losses. Bu the tower withstood the explosion (Figure 2.2.8). Six persons

were killed in the explosion and almost 1,500 were injured (WTC web site 2007). When

there can be an explosion at this proximity to the structure, blast resistant design must be

carried out and all the previously built structural members must be retrofitted to resist

blast loading.

Figure 2.2.8: WTC tower one after the 1993 explosion in the parking lot (WTC web site

2007)

23

Chapter 3

REHABILITAION FOR EARTHQUAKE

Although all the structures are designed for static loads such as live load, snow load, dead

load, etc. When dynamic loading is considered, ductility plays a pivotal role. It is of

paramount importance that the existing structures are retrofitted and the ductility be

improved in order to effectively dissipate the energy. This chapter evaluates some of the

effective rehabilitation techniques that improve the ductility and strength of the structure.

3.1. BEAM-COLUMN JOINT REHABILITATION There are many structures that are designed before the existence of the seismic codes.

Structures with insufficient shear reinforcement lead to a brittle behavior of the beam-

column joint. In addition to the above, due to the strong beam design the joint may be

subjected to high shear demand. It has been observer from recent earthquakes, 1995

Hanshin-Awaji (Kobe, Japan) and 1999 Kocaeli (Turkey), that the brittle failure of the

frame joints was the main cause that many structure collapsed (Mitchell et al. 1996,

Mugurama et al. 1995). Due to the large extent of this problem, it is very important to

develop an economical methodology to transfer the brittle failure of the joint to beam

flexure hinging mechanism, which is more ductile type of failure.

There were few studies conducted in view of improving the strength, stiffness, energy

dissipation and ductility of the beam-column joint. Kaun (1999) investigated a

rehabilitation procedure for damaged beam-column joints. Specimens were initially

damaged and then rehabilitated by epoxy resin injection technique. Then the specimens

were tested in cyclic loading test until failure. This method proved effective in improving

the strength and the energy dissipation of the joint. This method was not effective in

restoring the flexural stiffness of the beam and the shear stiffness of the joint.

Beres et al. (1992) proposed a rehabilitation scheme for beam-column joints. In this

method flat steel plates were added to the top and the bottom of the beam and these plates

were bolted to the continuous plate on the out side of the joint. Improvement in the

24

strengthened specimen was reported. When the results of the original specimen were

compare to the results of the specimen after retrofitting, large deterioration was observed

in the original specimen after the ultimate strength was reached. Ghobarah et al. (2001)

investigated a rehabilitating technique for RC beam-column joints. Deficient RC beam-

column joints were encased in a corrugated steel plate. This technique is very use full in

enhancing the shear strength of the joint.

Though several rehabilitation techniques are available for beam-column joint, the use of

fiber reinforced polymer (FRP) material in the rehabilitation of RC beam-column joint

offers several advantages (Ghobarah et al. 2001):

FRP rehabilitation is a fast process and is also applicable in tight locations.

FRP rehabilitation is non-disruptive to occupants and the functioning of the

building

FRP materials are resistant to corrosion

FRP materials are light compared to other rehabilitation materials.

FRP rehabilitation is simple and effective

3.1.1 Experimental Program

This section describes the work conducted by Ghobarah et al. (2001) which involved

experimental study to evaluate the effectiveness of FRP wrapping of RC beam-column

joint.

3.1.1.1 Specimen Description

Ghobarah et al. (2001) constructed a reinforced concrete column beam joint sample. The

height of the column is 3000mm (118.11 in) and has a cross section of 250 x 400 mm

(9.84 x 15.74 in). The length of the beam is 1750mm (68.89 in) from the face of the

column to the free end. The beam cross section is 250 x 400 mm (9.84 x 15.74 in). The

column has been reinforced with 6 M20 bars and 2 M15 bars in the longitudinal

direction. M10 bars are used for transverse reinforcement. M10 rectangular ties are

spaced at 200mm (7.87 in) center to center starting 80 mm above and below the beam.

The beam has been reinforced with 3 M20 bars on the top and 3 M20 bars on the bottom.

25

M10 bars are used for transverse reinforcement. The ties are spaced at 150 mm (5.9 in)

for 600 mm (23.62 in) and then spaced at 200 mm (7.87 in) for 1000 mm (39.37 in) and

ending at 75 mm (2.85 in) from the free end of the beam. The reinforcement arrangement

of specimen is shown in the Figure 3.1.1. (Ghobarah et al. 2001).

This specimen is designated as T1 and has been tested as a controlled specimen. The joint

is then repaired and rehabilitated with Glass fiber reinforced polymer (GFRP) and the

rehabilitated specimen is designated as T1R and the test were conducted on T1R.

Figure 3.1.1: Reinforcement details of the beam-column specimen (Ghobarah et al. 2001) 3.1.1.2 Material Properties

The Concrete used in the construction of the Beam-Column joint has a compressive

strength of 30.8 MPa on the day of the test. The concrete used in the repair of the joint

has a compressive strength of 38 MPa. The yield stress of the reinforcing steel is 425 and

26

454 MPa for M20 and M10 bars, respectively. The available FRP materials are carbon

and glass. Carbon fibers have higher strength and higher modulus, they are more suitable

for joint shear rehabilitation. However, carbon fibers are almost six times costlier that the

glass fibers. Considering this a GFRP has been used for this rehabilitation. The GFRP

laminate is a product of Fyfe Co. and is available commercially. Properties of the GFRP

are presented in the Table 3.1.1.

Table 3.1.1: GFRP material properties. (Ghobarah et al. 2001)

GFRP Ultimate Tensile Strength, MPa

Ultimate Elongation, %

Elastic Modulus, MPA

Thickness, mm

Bi-directional (in the 450 direction)

552 1.7-4.0 27579 1.1

After the T1 specimen is tested, the joint area had been cleaned of the fractured concrete

by an air hammer exposing all the bars. The joint was cleaned of all the debris and fine

particles using compressed air. Fresh concrete was then added to the joint. After four

weeks the surface of the column was cleaned and the edges are rounded and the fiber

laminate was wrapped. The specimen was tested after another four weeks. (Ghobarah et

al. 2001)

3.1.1.3 Experimental Test Set-Up

The beam-column is tested with column vertical position and the column supported at the

bottom and top. Restrainers are provided at the top and bottom of the column to take the

horizontal load. There is a 600 kN axial load in the column. The load is applied from the

top using a vertical jack mounted at the top of the column. The axial load is equivalent to

0.2 Agf’c where Ag is the gross area of the column section. The free end of the beam is

them applied with a cyclic load using a high capacity 1100 kN actuator of ±250mm

stroke. A schematic representation of the experimental setup is shown in the Figure 3.1.2.

3.1.1.4 Loading Sequence

Specimens were tested under reverse cyclic load applied at the beam tip. The selected

forces are indented to cause forces that simulate high level of inelastic deformations that

27

are normally experienced by the frame during an earthquake. The selected load history

consists of two phases. The first phase is load-controlled followed by a displacement-

controlled loading phase. (Ghobarah et al. 2001)

Figure 3.1.2: Experimental set up for the beam-column joint testing (Ghobarah et al. 2001) In the first phase of the loading two cycles of 15% of the estimated strength of the

specimen were applied on the specimen to check the test setup and to ensure sure all the

data acquisitions are functioning accurately. This was followed by two cycles of 22 kN

load (Concrete cracking load of the beam). These were followed by two cycles of loading

that causes initial yield of the bottom longitudinal steel bars in the beam. The yield

causing loads for the specimens T1 and T1R are 109 kN and 117 kN respectively. The

displacement at the initial displacement of the steel, δy, is used in the displacement

controlled phase. The displacement ductility factor, µ, is defined as the ratio between the

beam displacement to the displacement at the first yield (Ghobarah et al. 2001).

28

The second phase of loading starts after the steel yields in the first phase. This phase is

displacement controlled phase. The specimen was subjected to increasing displacement

starting from δ/δy= 2 using multiples of displacement previously recorded. Two cycles

were carried out at different ductility levels 2.0, 2.5, 3.0, etc., to verify the specimen

stability. The cyclic loading sequence is shown in the Figure 3.1.3.

Figure 3.1.3: Cyclic loading applied to the free end of the beam (Ghobarah et al. 2001) 3.1.1.5 Instrumentation

Different types of instruments were used to measure the displacement, load and strains.

Fourteen strain gauges were used to measure the strain in the steel reinforcement bars.

The location of the strain gauges is shown in the Figure 3.1.4. Two diagonal LVDT’s

(linear variable differential transducers) were used to measure the joint deformations as

shown in Figure 3.1.2. To monitor the relative deformation of the beam with respect to

the face of the column, two LVDT’s located above and below the beam were used. One

LVDT was used to measure the displacement of the beam tip. One LVDT is used to

measure the displacement of the top of the column. To measure the axial load applied. To

measure the axial load applied to the top and bottom of the column two load cells were

installed on the vertical jack and the cyclic load actuator.

29

Figure 3.1.4: Location of strain gauges on the reinforcement. (Ghobarah et al. 2001)

Figure 3.1.5: Rehabilitation scheme for the beam-column joint (Ghobarah et al. 2001)

30

3.1.2 Rehabilitation Scheme

The proposed rehabilitation scheme for the beam-column joint consists of wrapping the

joint with one layer of GFRP laminate in the form of a “U”. The free ends of the laminate

are tied together by steel plates and threaded steel rods driven through the joint as shown

in the Figure 3.1.5 and Figure 3.1.6. The height of the laminate is restricted to the joint

region and the potential presence of a slab will prevent extending the laminate above the

joint. The FRP is not extended to the beam because it will cause an increase in the

flexural strength and that will adversely affect the relative strength ratios of the connected

beam and column. This rehabilitation will enhance the shear capacity of the joint and

provide confinement. This method enhances the strength and ductility of the joint

(Ghobarah et al. 2001).

Figure 3.1.6: Beam-column joint with the FRP rehabilitation (Ghobarah et al. 2001)

31

3.1.3 Experimental Results

In the specimen T1, first crack was observed on the column face. A diagonal shear crack

in the form of an X-pattern was found on the column face before the yielding of

longitudinal beam steel. The joint shear capacity reduced as the beam tip displacement is

increased. These cracks extended to the back of the column at failure. A considerable

degradation in strength occurred at ductility factor of 2, test was terminated at ductility

factor 2.5 when the load carrying capacity dropped to 30% of the maximum load.

(Ghobarah et al. 2001)

The T1R specimen which had been cracked from the initial testing is tested now. At a

ductility factor of 2.5 there is a slight delamination of the fiber as a slight finger tap on

the fiber revealed a hollow sound. The delamination area increased until it covered the

whole column face. The presence of the steel jacket prevented the premature failure of

the joint. The joint has been held in place by the jacket even after fiber delamination. Due

to the intentional under design of the fiber, failure of the fabric dint until a ductility factor

of 4. It started in the diagonal direction as a tear while pushing up, further tears extended

in the other diagonal from the extremities of the original tear. When the load is reversed,

the fiber material completely separated into two pieces revealing the failure of the

concrete underneath. The crack patterns are shown in Figure 3.1.7 (Ghobarah et al.

2001).

32

Figure 3.1.7: FRP laminate failure of specimen T1R (Ghobarah et al. 2001)

3.1.4 Summary

The original sample with no retrofitting in the joint region showed rapid strength

deterioration once the longitudinal steel of the beam started yielding. This was due to the

brittle shear failure of the joint. The rehabilitated specimen showed superior energy

dissipation characteristics compared to the original specimen. Use of higher yield and

strain hardening values for steel reinforcement than the nominal design values may result

in greater strength in the beam flexural capacity. This may cause excessive shear stresses

in the joint. This should be taken into account when seismic rehabilitation of beam-

column joint is carried out.

33

3.2 SQUARE HIGH-STRENGTH CONCRETE COLUMNS IN FRP STAY-IN-

PLACE FORMWORK

High strength concrete is superior to normal-strength concrete in terms of strength and

performance. The use of high strength concrete in buildings and bridges has been

increasing over the last two decades. However, the use of high strength concrete in

seismically active zones has been limited because of the fact that the increase in the

strength and performance has been achieved by compromising deformability. In a seismic

zone inelastic deformation and efficient energy dissipation are required to resist seismic

forces. High strength concrete structural elements exhibit brittle behavior at failure which

is not recommended for seismic loading.

Inelastic deformation of concrete can be improved by providing better confinement.

Significant lateral drift can be achieved without significant strength deterioration by

providing properly designed transverse reinforcement. When conventional steel ties,

hoops, overlapping hoops or spirals are used for high strength concrete the confinement

requirements are very high and are not acceptable. Using conventional confinements with

high strength concrete will create cage congestion and concrete filling problems. FRP

laminate offer a alternative and effective method to provide confinement to high strength

concrete (Ozbakkalogu et al. 2007).

FRP used as stay-in-place formwork have several advantages:

Light and provide effective framework with superior handling characteristics

Effective and has the ability to develop high lateral confinement pressures

Acts as a protective shell against corrosion, weathering and chemical attack

3.2.1 Experimental Procedure

Considering the confinement of the concrete column it has been long proven that spiral

confinement works effectively compared to a rectangular confinement. Similarly it has

been reported that FRP jackets are effective in circular columns than in square columns

(Mirmiran et al. 1998b, Pessiki et al. 2001). This can be explained by hoop tension that

34

develops in circular columns and helps in maintaining a uniform passive confinement

force. Whereas in square columns high confinement forces develop at the corners

As the corner radius (R) is increased, it promotes hoop tension in the fiber and improves

the effectiveness of the confinement. The cross-sectional size (D) affects the flexural

rigidity of the FRP fiber between the corners and thereby affecting confinement

efficiency. These two parameters can be expressed as ration R/D. Two R/D ratios were

used for this experimental procedure 1/16 and 1/34.

Internal crossties are required in conventionally reinforced columns; they improve the

distribution of lateral loads and restrict the concrete from expanding and improving the

confinement action. A similar concept is used to provide crossties in the FRP stay-in-

place formwork (Ozbakkalogu et al. 2007).

3.2.2 Test Specimens

A total of six specimens were prepared and each of them has a 270 mm square cross

section and a 1720 mm cantilever height. The shear span of each column is 2000mm

measured to the point of load application, which is 280mm above the beam. These

specimens represent the lower half of the first story building column. The specimen

configuration is shown in Figure 3.2.1.

35

Figure 3.2.1: Geometry of the columns specimen used in the testing (Ozbakkalogu et al.

2007)

High strength concrete, with cylinder strengths 75 MPa and 90 MPa, was used to cast the

columns. A clear cover of 20mm is provided, measuring from the face of the column to

the outside of the longitudinal reinforcement. Three different sets of reinforcement

arrangements are used 4-bar, 8-bar, and 12-bar. For the latter two arrangements number

15 deformed steel bars with yield strength of 500 MPa are used and for the arrangement

with 4-bars number 20 bars are used with yield strength of 476 MPa (Ozbakkalogu et al.

2007).

Following are the description of the arrangement of the reinforcement and the FRP

formwork and crossties for different specimens.

RS-1

Specimen RS-1 has four number 20 bars for reinforcement and has five plies of FRP in

the casing. The corner radius for this specimen is 45mm.

36

Figure 3.2.2: Reinforcement arrangement used in specimen RS-1 (Ozbakkalogu et al.

2007)

RS-2

Specimen RS-2 has eight number 15 bars for reinforcement as shown in the Figure 3.2.3

and has five plies of FRP in the casing. The corner radius for this specimen is 45mm.

This specimen also has FRP crossties in both the cross-sectional direction.


2007)

RS-3

Specimen RS-3 has twelve number 15 bars for reinforcement as shown in the Figure

3.2.4 and has five plies of FRP in the casing. The corner radius for this specimen is

45mm. This specimen has two crossties in each of the cross-sectional direction separated

by a distance of 68mm, ¼ of the column dimension.

37


2007)

RS-4


and has three plies of FRP in the casing. The corner radius for this specimen is 45mm.


2007)

RS-5


and has two plies of FRP in the casing. The corner radius for this specimen is 45mm.

38


2007)

RS-6


and has three plies of FRP in the casing. The corner radius for this specimen is 8mm.

This specimen also has FRP crossties in both the cross-sectional direction.


2007)

3.2.3 Material Properties

3.2.3.1 Carbon FRP composite

Carbon fiber is preferred over glass fiber because of the fact that carbon fibers exhibit

higher elastic modulus and tensile strength of the material, which is more compatible

with high modulus and high strength concrete. The normal thickness of the carbon fiber

39

sheet is 0.165mm/ply and this has been increased to 0.8mm/ply after the laminate has

been impregnated with epoxy resin. The same FRP composite has been used for all the

casings and the fiber is allied in the transverse direction (Ozbakkalogu et al. 2007).

The manufacturing process for the FRP casings was done by wrapping impregnated FRP

sheets around a wooden template. For the columns with 45mm corner radius, PVC tubes

were used for rounding the corners. For the columns with 8mm corner radius wooden

quarter rounds were used for rounding the corner as shown in the Figure 3.2.8.

Figure 3.2.8: Wooden templates used to make the FRP casings (Ozbakkalogu et al.

2007)

The wrapping of the casings were done layer by layer. Each layer had a overlap of

100mm in the direction of the fiber to ensure proper bonding. There is no overlap of

adjacent layers along the height of the column. Hand lay-up technique used here resulted

in good quality casings, an advanced automated method like centrifugal casting, filament

winding or pultrusion may offer better quality.

40

The crossties were also made out of the same FRP fiber that is used in the preparation of

the casings. Crossties were prepared by wrapping the FRP around a low tensile strength

phenolic bar. The phenolic bar doesn’t contribute to the strength of the crosstie, but

simple act as a template (Ozbakkalogu et al. 2007).

Two different fiber contents were used in preparing the crossties. In the specimen RS-2

and RS-3, 40mm strips of fiber are used to make the crossties. The crossties were mainly

provided to resist against lateral expansion of concrete. The contribution of crossties to

the total fiber cross-section area is very small.

In the specimens RS-5 the number of plies in the casing was reduced by one layer and the

reduced amount of fiber was used to make the crossties. The thickness of the crosstie for

RS-5 is 136mm. The total area of fiber for the specimen RS-5 with two plies in casing

and crossties is equal to that of specimens RS-4 and RS-6 with 3 plies in casing and no

crossties (Ozbakkalogu et al. 2007).

3.2.3.2 Concrete Properties

Two concrete mixes were used to make the six specimens. First three specimens RS-1,

RS-2 and RS-3 are made from 10SF cement and crushed lime stone with a maximum size

of 10 mm with a water cement ratio of 0.22. The target strength for this concrete is 90

MPa. The obtained concrete strength is 90.1 MPa. The second mix used to make the

specimens RS-4, RS-5 and RS-6 is a mixture of 10SF cement and crushed lime stone

with a maximum size of 10 mm. The water cement ratio used is 0.26 and the target

strength for this concrete is 75 MPa. Obtained concrete strength is 75.2 MPa. All

columns were cast vertically and vibrated thoroughly. The strength of the specimens was

monitored timely by testing cylinders. Tests were conducted when the average strength

reached the target strength.

3.2.3.3 Steel Reinforcement

Canadian standard No. 15 and No. 20 deformed bars with a nominal diameter of 16 mm

and 19.5 mm are used as longitudinal reinforcement with yield strength of 500 and 476

41

MPa. Mechanical spices are used to splice longitudinal bars at 500mm and 900mm from

the footing interface.

Table 3.2.1: FRP casing and crosstie details (Ozbakkalogu et al. 2007)

Column Shear Span

(mm)

f’c

(MPa)

Number

of plies R/D

Crosstie area

(mm2)

RS-1 2,000 90.1 5 1/6 0

RS-2 2,000 90.1 5 1/6 6.6

RS-3 2,000 90.1 5 1/6 2 x 6.6

RS-4 2,000 75.2 3 1/6 0

RS-5 2,000 75.2 2 1/6 22.3

RS-6 2,000 75.2 3 1/34 0

Table 3.2.2: Reinforcement details (Ozbakkalogu et al. 2007)

Column

fy

MPa

Reinforcement

arrangement

Longitudinal

reinforcement

ratio

RS-1 476 4 – No. 20 1.68

RS-2 500 8 – No. 15 2.24

RS-3 500 12 – No. 15 3.36

RS-4 500 8 – No. 15 2.24

RS-5 500 8 – No. 15 2.24

RS-6 500 8 – No. 15 2.20

3.2.4 Experimental Test Setup

The columns were fitted with linear variable displacement transducers and strain gauges

to measure the horizontal displacement, anchorage slip, horizontal and transverse strains

and rotation of plastic hinge. All equipment was connected to a microcomputer and data

acquisition system for recording and analyzing data.

42

Each column specimen was tested under incrementally increasing lateral deformation

reversals, simulating seismic load and a constant axial load. Two 1,000 KN capacity

computer servo-controlled MTS hydraulic actuators are placed vertically on two sides of

the column to simulate the axial load applied on the column due the stories above the

column. This column represents the first story column of a multi story building. The

specimens were tested with 30 or 34% of the total concentric capacity, Po, computed

using the following equation.

yssgco fAAAfP +−= )(85.0 '

Where, fc’ = compressive capacity of concrete

Ag = Gross area of the column section

As = Total area of longitudinal steel

The specimens were subjected to incrementally increasing lateral reversal loads. Three

cycles were applied at each of the drift ratio starting at 0.5% and increasing to 1%, 2%,

3%, etc., in the deformation control mode of the horizontal actuator. The lateral load was

applied on the specimen until it looses a large fraction of its maximum lateral load

resisting capacity. The rate of lateral loading was low and the typical duration of the test

is 4-5hrs, depending upon the deformability of the specimen (Ozbakkalogu et al. 2007).

3.2.5 Test Results

Specimens RS-1, RS-2, RS-3, RS-4 and RS-5 behaved almost in the same manner until

2% lateral drift ratio. Specimen RS-6 with small corner radius started demonstrating

signs of distress during the first cycle of 2% lateral drift. All the column specimens with

well rounded corners showed no visual damage until the end of the 2% lateral drift ratio

cycles. At 3% lateral drift ratio localized color change was observed on the FRP on the

column specimens RS-1, RS-4 and RS-5, which have either reduced plies or have no

crossties. This indicated in these columns specimens, there is separation of the FRP

laminate from the concrete due to crushing of concrete. A similar discoloration has been

observed in the column specimens RS-2 and RS-3 which have five plies and also the

crossties at 3% lateral drift ratio. The region discoloration increased as the lateral

43

displacement was increased until it is 540mm from the column-footing interface which is

twice the dimension of the column cross-section.

Specimen RS-6 started to expand into a circular shape beyond 2% lateral drift and at 4%

lateral drift ratio fiber rupture began. These phenomena are not prominent in the columns

with well rounded corners and crossties (Ozbakkalogu et al. 2007).

Specimens RS-1, RS-2 and RS-3 all have the same number of plies in the casing. RS-1

has no crossties, RS-2 has single crossties in each of the cross-sectional directions and

RS-3 has two crossties in each of the cross-sectional directions. Fiber rupture occurred in

RS-1 in the third cycle of 8% lateral drift ratio. Similarly in specimen RS-3 fiber rupture

occurred at 12% lateral drift ratio and in RS-2 before the fiber reached failure the test was

stop due to out of plane deformations after the completion of third cycle of 9% lateral

drift ratio.

Specimens RS-4 and RS-5 have the same amount of fiber per cross-section only

difference being RS-4 has three plies in the casing and RS-5 has two plies in the casing

with the rest of the fiber being used for crossties. Both these columns exhibited similar

behavior, with RS-4 failing at the first cycle of 7% lateral drift ratio and RS-5 failing at

third cycle of 6% lateral drift ratio (Ozbakkalogu et al. 2007).

3.2.6 Deformation of HSC Columns Confined by FRP Casing

The deformation of high strength concrete is a major concern due to the fact that high

strength concrete is brittle in nature, especially in the presence of axial load.

Deformability is a major factor in the seismic design of structure and the major drawback

of high strength concrete over normal concrete in terms of seismic design. The brittle

nature can be improved by the use of FRP casing and crossties.

When column with five plies of FRP and with an R/D ratio of 1/6 tested it developed 8%

lateral drift ratio. When the column with five plies of FRP in the casing, two crossties in

each direction of the cross-section and with R/D ratio of 1/6 it developed 11% lateral drift

44

ratio. When three plies are used instead of five plies for the casing and with the same R/D

ratio of 1/6 the maximum lateral drift ratio has decreased to 6%. However when the ratio

of corner radius to column dimension was minimized to R/D = 1/34 there is a substantial

reduction in the lateral drift capacity of the column and it developed only 2% lateral drift

ratio (Ozbakkalogu et al. 2007).

3.2.7 Summary

Due to the brittle behavior of the high strength concrete, the confinement requirements

are very high to resist earthquake loading and behave in a ductile fashion. FRP stay-in-

place confinement for high strength concrete will give better results compared to the

conventional steel confinement, covering the whole column. The ratio of the corner

radius to the cross-section dimension of the column (R/D) has a major effect on the

effectiveness of the FRP casing. Providing a rounded corner will prevent the premature

failure of the fiber due to high stresses at the corners. The use of FRP crossties will

improve the efficiency of the FRP confinement in the similar way over lapping hoops and

crossties do in the conventional steel confinements. The concept of integrated crossties in

FRP casing has proven to be effective.

3.3 REHABILITATION OF COLUMNS WITH REINFORCED CONCRETE JACKETS

An RC element can be repaired by attempting to restore the original strength and stiffness

of an RC element that is either damaged or deteriorated. There is a distinction between

cosmetic repair and structural repair in his repair of RC columns. If the loss in strength is

lower than 10%, cosmetic repair is considered and if the decrease in strength is above that

value, structural repair is considered. Repairing a damaged RC element just by replacing

some of the original materials does not restore the characteristics of the original element.

Therefore, this method is acceptable only in the case of cosmetic repair (Julio et al.

2003).

3.3.1 Added Longitudinal Reinforcement

One of the advantages of RC jacketing strengthening of columns is that the increase in

stiffness of the structure is distributed uniformly distributed as compared to addition of

45

shear walls or bracing. While jacketing an existing column the added longitudinal

reinforcement bars must be anchored to the foundation. In this method usually it is

necessary to execute a new foundation or at least strengthen the existing one (Julio et al.

2003).

Although several commercial products, effective in bonding the newly added longitudinal

steel to the foundation, are available care must be taken while executing this process.

Details must be considered to ensure proper bonding.

Julio (2001) conducted several test on RC columns strengthened by jacketing. The newly

added reinforcement for the jacketing was anchored to the footing using a commercially

available two component resin. Then the samples were submitted to monotonic tests, with

a constant axial load and increasing shear force and bending moments. The failure of the

longitudinal steel bars of the existing column and slippage of the newly added bars was

observed as shown in the Figure 3.3.1.

Figure 3.3.1: Failure of the steel bars of the column and slippage of the steel bars of the

added jacketing (Julio et al. 2003)

Pull out tests were conducted and it was concluded that slippage of the longitudinal bars

in the jacketing is the main reason of failure. This occurred due to the fact that the holes

46

drilled were not cleaned properly. Cleaning the driller holes with a vacuum cleaner is

enough from changing the failure from a slippage failure to tension rupture.

3.3.2 Slab Crossing

To retain the structural integrity the longitudinal bars added to the structure must go

through the slab and maintain continuity. For the longitudinal bars to pass through holes

must be drilled through the slab. Alcocer et al. (1993) indicates that the use of column

distributed reinforcement is better that column bundles to reduce the possibility of bond

damage. But in the case of a column beam system there is always a possibility of

interrupting the middle bars, so longitudinal reinforcement can be provided in the corners

so as to avoid the interruption.

3.3.3 Interface Surface Preparation

It is important to prepare the interface to achieve good bond between the original column

and the added concrete, so that the whole unit acts monolithically. There are several

methods in use to improve the surface roughness of the original columns. Following are

some of these techniques (Julio et al. 2003).

3.3.3.1 Methods to Increasing Surface Roughness

Several methods are used in increasing the roughness of the original column and thereby

increasing the bond between the original column and jacket. Following are some of the

methods to improve the surface roughness (Julio et al. 2003):

1. Hand Clipping

2. Sand Blasting

3. Jack Hammering

4. Electric hammering

5. Water Demolition

6. Iron Brushing

Roughness of the original column surface is a important factor for the strength of the

column but that has been quantified.

47

Samples made with different surface roughening methods were studied for the bond

strength of the original concrete and the jacket. One important conclusion is that

pneumatic hammering causes micro cracks in the substrate. This is one of the methods

predominantly used to increase the surface roughness. This method has to be avoided

since it has been proved that the mechanical action of the hammer weakens the joint.

Julio et al. (2003) conducted several experiments to study the influence of the interface

between concretes of different ages on the strength of the joint. The surface prepared by

different techniques on left as cast specimens were tested. Slant shear test and pull-off

test were conducted on specimen interface prepared by sand blasting (Figure 3.3.2),

prepared by electric hammering and treated with iron brushing. It has been observer that

sand blasting is the most effective of the methods considered.

Figure 3.3.2: Specimens prepared by sand-blasting (Julio et al. 2003)

3.3.3.2 Surface Pre-wetting

The question of pre-wetting the surface is inconclusive. The AASHTO-AGB-ARTBA

joint committee recommended that the original surface must be dry before the new layer

of concrete is cast. The Canadian Standards Association standard A23.1 recommends the

original concrete surface be wetted for at least 24 hours before the cast of the new

concrete.

48

A critical amount of moisture must be maintained in the substrate to achieve maximum

strength. An excess amount of moisture in the substrate will close the pores and prevent

the absorption of the repairing material. An excessively dry substrate can absorb too

much water from the repairing material causing excessive shrinkage in the repairing

material. A saturated substrate with a dry surface is considered the best solution (Julio et

al. 2003).

3.3.3.3 Application of Bonding Agents

There are some published works on the effectiveness of the adhesion between repair

materials and concrete substrates with bonding agents. The conclusion reached by

different authors is not always the same. Due to the enormous variety of parameters

influencing the interference strength the results obtained are not comparable.

Julio et al. (2003) conducted slant shear tests and pull-out test on specimens with

different interfaces. Specimens were prepared by different techniques of surface roughing

like sand blasting, electric hammering and iron brushing and two component epoxy resin

was used as a bonding agent as shown in Figure 3.3.3 and the specimens were tested. The

value of the shear and tensile strength of the specimen reduced when epoxy resin was

used with sand blasting, while shear and tensile strength increased when epoxy resin is

used with the other techniques.

Figure 3.3.3: Application of epoxy resin on specimen (Julio et al. 2003)

49

3.3.3.4 Addition of Steel Connectors

Steel connectors (Figure 3.3.4) play an important role in case of precast RC beams with

in situ cast slabs. Julio (2001) conducted push-off on specimens to analyze the effect of

steel connectors on the interface strength. Julio conducted tests on seven different

specimens with seven different surface penetrations were considered. It was observed that

the addition of the steel connectors did not increase the debonding strength but increased

the longitudinal shear strength considering the slippage. Two commercial products were

used to anchor the steel connectors to the concrete. The fact that the steel connectors were

added after the concrete was cast by drilling holes dint reduces the joint strength.

Figure 3.3.4: Steel connectors epoxy bonded on push off specimens (Julio et al. 2003)

3.3.3.5 Testing of Different Methods

Julio (2001) conducted monolithic and cyclic tests on jacketed undamaged RC specimens

prepared by different methods. He considered six different specimens with different

interface treatments: a non strengthened column, a monolithic model, a column without

interface surface preparation, a column strengthened with interface surface prepared by

50

sand blasting, a column with same roughness treatment as the previous specimen with

added steel connectors and a model where non-adhesive between the original column and

the added jacket was artificially induced. Except for the last specimen all other specimens

acted monolithically when subjected to monotonic test and cyclic tests. The major

conclusion of the author was contradictory to the current practice in some countries.

There is no need to improve the interface surface or use any bonding agent to improve the

joint strength for undamaged RC columns.

3.3.4 Spacing of Added Stirrups

The coefficient of monolithic behaviors of the RC columns strengthened by jacketing has

been evaluated and it has been concluded that a higher percentage of transverse

reinforcement gives better confinement to the strengthened column to achieve better

monolithic performance. It has been recommended that half the value of the transverse

reinforcement spacing in the original column be adopted for the transverse reinforcement

spacing on the strengthening jacket.

3.3.5 Temporary Shoring of the Structure

One of the major factors in strengthening of a column is how it has been done.

Strengthening of a loaded column is lot different from strengthening of an unloaded

column. When a loaded column is strengthened the original column is resistant to these

and to the loads already applied. In the second case where an unloaded column has been

strengthened the composite structure of the original column and the RC jacket will act as

a unit and resist the total load. When a combined action of the original column and the

RC jacket is desired, the load on the column must be temporarily transferred to a

temporary shoring so as to strengthen the column as an unloaded column (Julio et al.

2003).

3.3.6 Properties of Added Concrete

Normally the added concrete has aggregate of maximum dimension of 2mm because of

the lack of space in the jacket. For the same reason self compacting concrete (SCC) is

used for jacketing. This is due to the diminished thickness occupied by the concrete due

51

to the volume occupied by the steel. For this reason high strength concrete (HSC) is a

good option while strengthening RC columns with concrete jackets. High strength

concrete is obtained by silica fume addition. Since the substrate of the original concrete is

much older than the new concrete used or the jacketing, it is recommended to use

concrete with less shrinkage.

Julio (2001) conducted test studying the effect of the strength of the joint when different

strength concrete were used in the jacketing procedure. It has been concluded that the

strength of the joint increased as the nominal strength of the concrete used for the

jacketing increased. When high performance concretes (HPC) were used the failure mode

shifted from interface rupture to monolithic failure. Hence HPC is preferred option for

RC column jacketing.

Figure 3.3.5: Concrete casting of the jacket (Julio et al. 2003)

3.3.7 Structural Behavior

A significant increase in the strength and ductility of the columns can be achieved with

this rehabilitation. This method will also alter the overall behavior of the building.

52

Alcocer et al. (1990) conducted several experiments on RC columns strengthened by

jacketing; the specimens were tested by applying bidirectional cyclic loading. The

authors have concluded that jacketing of RC columns can change a system from weak

column strong beam to strong column weak beam scenario.

3.3.7.1 Effect of Damage on Structural Behavior

There is a lot of difference in strengthening a healthy column and heavily damaged

column. Alcocer et al. (1993) conducted experimental test on jacketed RC frames, by

jacketing the most damaged elements, joints and columns, the strength values obtained in

the test at 2% drift was 63% of the undamaged specimen and the stiffness values

obtained at 0.5% drift is 52% of the values obtained for a undamaged specimen.

Rodriguez et al. (1994) performed tests to see the effect of jacketing on damaged and

undamaged specimens to investigate the increase of strength, stiffness and ductility of the

specimens under seismic load conditions. The author built specimens as per the 1950s

New Zealand code and concluded that these specimens have low ductility and were not

suitable for seismic loading. These specimens were jacketed and tested. The author

concluded that the ductility of the jacketed specimens was almost three times the original

as build columns. The author also mentioned that the extent of damage and the

reinforcement details had very little influence on the overall seismic performance.

3.3.8 Summary

RC jacketing technique improves the strength and stiffness of the column. Unlike other

methods this method leads to a uniform increase in the strength and stiffness of the

member. This method does not need any specialized workmanship and this makes RC

jacketing a valuable choice for structural rehabilitation. In this method attention must be

paid to the following aspects.

• The use of hand clipping, jack hammering, electric hammering to remove the

damaged concrete causes micro cracks in the substrate, so sand blasting or water

demolition techniques must be used.

53

• For undamaged columns no additional interface surface preparation is required

except for short columns. Where ever interface surface preparation is necessary

sand blasting or water demolition techniques must be used.

• Use of bonding agent – a two-component epoxy resin is most commonly used.

When trying to improve the roughness of the interface surface an effective

method like sandblast is sufficient. When epoxy resin is used along with sand

blasting technique reduction in the strength is observer, hence this must be

avoided.

• Application of steel connectors – these must be used only in case of short

columns to improve the level of strength and stiffness under cyclic loadings.

• Temporary shoring – care must be takes in such a way that the concrete jackets

resists part of the total load rather than part of the incremental loads on the

column.

• Continuity of the longitudinal bars through the slab is must to retain the integrity

of the structure. So holes must be drilled to allow the longitudinal bars to pass

through the slab.

• The longitudinal reinforcement must be spread uniformly on all sides of the

column. When ever this is not possible care must be taken to avoid excessive

bundling at the corners.

• Added stirrups – stirrups in the jacketing must be placed with a spacing of half the

spacing that is used in the original column’s stirrups.

• Added concrete – self-compacting, high-strength and high-durability concrete,

non shrinkage concrete must be used.

3.4 SEISMIC REHABILITATION BY ADDING SHEAR WALLS

After the 1995 Dinar and 1996 Adana-Ceyhan earthquakes that measured 6.0 and 6.2

magnitude, several buildings are damaged in the towns of Dinar and Ceyhan in Turkey.

Under the guidance of the Earthquake Engineering Research Center of the Middle East

Technical University a total of 130 moderately damaged structures are rehabilitated.

54

The following are the main factors leading to the building damages in Dinar and Ceyhan

(Sucuoglu et al. 2004):

There is no closely spaced confinement reinforcement at beam column joints.

Beams were generally stronger than columns in all stories.

The measured concrete strength is usually less than 15 MPa.

Plain reinforcement with yield strength of 220 MPa bars is used in all buildings.

Sufficient anchorage lengths were not provided for longitudinal reinforcement.

Transverse beam and column are not providing enough confinement.

3.4.1 Rehabilitation Method

Considering the weaknesses of the structure, damaged condition, constrain of completion

of rehabilitation in limited time, a simple technique of rehabilitation has been used for all

the buildings that are damaged. The system used in the rehabilitation of the buildings is to

add concrete shear walls in the existing structure. The newly added shear wall system

will act as the primary system for the seismic loads while the existing system will be the

secondary system. The existing frame work will still be the primary load bearing system

for the gravity loads (Sucuoglu et al. 2004).

3.4.2 Design criteria for seismic design

Since the newly added shear walls must act as the primary system resisting the seismic

loads, the design criteria adopted must ensure that the shear walls added sufficient

stiffness and strength rehabilitated system. Thus, the new walls and the connections were

to take the maximum lateral load and reduce the lateral deformations to acceptable level.

The following are the design criteria adopted (Sucuoglu et al. 2004):

Seismic design was conducted as per the Turkish code.

The material properties that are obtained in the in-situ testing were used in the

design.

The newly added shear walls have to resist at least 70% of the lateral load.

The axial loads in the columns developed both due to the seismic loads and

gravity loads combined must be less than 50% of the axial load capacity of the

column. Otherwise columns are strengthened to increase the axial load capacity.

55

In both the orthogonal directions, the ratio of the cross sectional area of the shear

walls and the total floor area has to be more than 0.002.

Confinement reinforcement was provided at the edges of the shear walls adjacent

to the columns.

The continuity of the vertical bars in the shear walls is maintained by providing

vertical dowelled bars.

To resist the over turning moment, new foundations were provided under the

newly added shear walls.

3.4.3 Four Story Building in Dinar

The four story structure that is shown in the Figure 3.4.1 has been damaged moderately in

the Dinar earthquake. It was rehabilitated by adding shear walls that extend all the way

from the ground floor to the top floor. The layout of the building and the location of the

shear walls have been shown in the Figure 3.4.2.

Figure 3.4.1: A four story building that’s been rehabilitated in Dinar (Sucuoglu et al.

2004)

56

The building was a reinforced concrete frame with concrete slabs and individual footings

connected by foundation tie beams in both directions. The typical floor area is 310 m2

and the floor height is 3.80 m for the ground floor and 3.50 m for all floors above the

ground floor. The most common size of column cross section is 25 cm by 60 cm and all

the beams are 25 cm by 70 cm. There is a 1m cantilever on the front and back of the

building starting from the first floor. There are fewer walls in the ground floor and that

created a weakness.

During the 1 October 1995 Dinar earthquake, all of the 23 columns had been damaged in

the ground floor, two of the columns had been severely damaged with shear failure on

both ends, three were damaged moderately and rest 18 columns were damaged lightly. In

the ground floor all the brick partitions were heavily damaged. In the first floor out of 37

columns two columns were damaged severely and three columns were damaged

moderately and one was damaged lightly. There is visible damage in the brick partitions

even in the upper floors. There was not much damage for the beams, clearly showing that

beams were stronger that the columns. The transverse reinforcement provided in the

columns and beams had a spacing of 20-25cm. The concrete core samples that have been

taken from the building revealed a mean concrete strength of 12 MPa (Sucuoglu et al.

2004).

57

Figure 3.4.2: The Rehabilitated scheme of the four stories structure in Dinar, ground

floor plan (Sucuoglu et al. 2004)

In the rehabilitation process of this building two U-shaped shear walls were added to the

building as shown in the Figure 3.4.2 and the corner columns were strengthened. Column

6C was severely damaged. Since this column was to remain in one of the shear walls,

concrete was completely removed from this column and has been recast along with the

shear wall. Column 6B was also damaged severely. This column was jacketed with a

concrete cover. The shear wall ration in the rehabilitated building in the x and y

directions is 0.0021 and 0.0032 respectively.

3.4.4 Eight Story Building in Ceyhan

This building (Figure 3.4.3) has been damaged moderately in the June 25, 1998 Adana-

Ceyhan earthquake. Damage was mainly observed in beams in the first five levels,

columns and the U shaped shear wall around the elevator in the first floor. As a part of

the rehabilitation process shear walls were added to the structure as shown in the Figure

58

3.4.4. The areas denoted with darker shaded regions are the newly added shear walls

(Sucuoglu et al. 2004).

The structural system of the existing structure was almost symmetrical in the long

direction. This structure had a two way continuous footing at a depth of 1m from the

ground. The beam sizes were 20 mm by 60 mm and the thickness of the concrete slab is

14 cm. The column sizes are 25 mm x 50 mm and 25 mm x 70 mm. The floor area of the

ground floor is 195 m2 and the floor area of the floors above it is 227 m2. The floor height

all through the building is 3 m for all the floors.

Figure 3.4.3: Eight story damaged building in Ceyhan (Sucuoglu et al. 2004) During the Adana-Ceyhan earthquake four of the 24 columns in the first floor in this

building were lightly damaged and the other columns are not damaged. Out of the three

existing concrete wall segments two were lightly damaged and one is moderately

damaged. Most of the 55 beams and 36 infill walls in the ground floor of the structure

were moderately damaged. Similar beam and infill wall damages patters were observed

in the upper floors all the way to the fifth floor.

59

Majority of the connections satisfied a strong column weak beam failure. The

longitudinal reinforcement of the columns was between 1 - 2%. The transverse

reinforcement provided in the column had a spacing of 20mm. All the beams in the

structure were reinforced lightly. A favorable beam mechanism was observed by

considering the beam damage distribution over the five stories, but the infill wall

damages was an indication of the lack of lateral stiffness and strength. The core concrete

samples were collected from walls and beams showed a mean concrete strength of 14

MPa (Sucuoglu et al. 2004).

During the seismic rehabilitation of the structure four infill shear walls were added to the

structure as shown in the Figure 3.4.4. The wall ratio was increased to 0.0020 in the x

direction and 0.0027 in the y direction.

Figure 3.4.4: Rehabilitation scheme of the eight stories building in Ceyhan, ground floor plan (Sucuoglu et al. 2004) 3.4.5 Aftershock Test in Dinar

A magnitude 4.6 after shock occurred on 4 April, 1998 on the Dinar fault. Since all the

rehabilitation work on 35 buildings is completed, this after shock has been considered as

60

a live performance evaluation for the rehabilitated building for seismic forces. The effects

of this after shock is considerably minimal on the 35 rehabilitated buildings. This after

shock caused hairline cracks at the infill walls and existing structure interface. Some of

the buildings also showed hairline cracking of masonry infill wall. The four story

building that has been discussed showed minor shear cracking on the masonry piers

between the windows of the second story which snapped the cantilevering facade. Hair

line cracks at the roots of the first floor beams connecting to the shear walls are observed

(Sucuoglu et al. 2004).

3.5 PASSIVE SEISMIC PROTECTION IN STRUCTURAL REHABILITATION

Passive seismic protection system is commonly referred as a seismic protection system.

A set of devices, through their action increases the seismic capacity of the structure.

These systems will reduce the lateral deformation by modify the dynamic global behavior

of the structure or by increasing the energy dissipations capacity of the assemblies, there

by reducing the forces and deformations (Guerreiro et al. 2006).

There are three types of passive seismic protection systems

1. Passive seismic protection systems

2. Active seismic protections systems

3. Semi-Active seismic protection system

The main difference is that passive seismic protection systems do not need any energy

supply for their normal behavior. Active and Semi-Active seismic protection devices

need energy supply for their normal behavior. Base isolator or the use of energy

dissipation devices are the most important passive seismic protection systems. These

technologies are currently used in seismic rehabilitation of old structures as well as in

new structures (Guerreiro et al. 2006).

Traditional seismic rehabilitation techniques try to improve the capacity of the structure

to resist lateral loads by adding reinforced concrete shear walls or rigid frame structures.

The modification of the global stiffness of the structure, in most cases, increases the

61

natural frequency thereby increasing the seismic demand. In order to avoid this situation

it is important to rehabilitate the structure with out increasing the seismic demand. The

use of base isolator is an effective method of improving the seismic resistance of the

structure without increasing the seismic demand of the structure.

3.5.1 Base Isolation

Base isolator decouples the base horizontal movement from the horizontal movement of

the structure. This is obtained by introducing a horizontal layer with high horizontal

flexibility between the super structure and the foundation. By reducing the horizontal

stiffness of the structure the fundamental frequency of the structure has been reduced to a

lower level compared to a fixed base system. By using a base isolation system we are

reducing the fundamental frequency of the structure there by reducing the seismic force

demand of the super-structure. Since the isolation system has a high horizontal flexibility

the deformations will be large and care should be taken to accommodate high

deformations. Deformations are concentrated at the isolation level, where the base

isolation system is located and is designed to absorb them. The use of base isolators has

two advantages. It reduces the inter story drift and the floor acceleration (Guerreiro et al.

2006).

The base isolation can be achieved by introducing a special device at the isolation level,

creating the flexible layer of decoupling. The isolators must have low horizontal stiffness,

capacity to support vertical forces, horizontal restoring force to re-centre the structure

after the motion, capacity to dissipate energy.

The following are the most common base isolation systems

1) High-damping rubber bearings (HDBR)

2) Lead rubber bearings (LRB)

3) Friction pendulum systems (FPS)

62

Figure 3.5.1: Different types of base-isolation systems (Guerreiro et al. 2006)

Figure 3.5.2: Example of retro-fitting a RC building with base isolation (Guerreiro et al.

2006)

The major advantage of base isolation technique is that the seismic demand of the

structure can be reduced to the same magnitude as the available capacity of the original

structure, even when the available capacity of the structure is rather low. This can be

achieved by lowering the natural frequency and effectively maintaining the energy

dissipation. By implementing this technique the horizontal forces on the structure are

significantly reduced compared to rigid base structure there by reducing the retrofitting

requirement. Base isolation is a modern technique which can be predominantly used in

retrofitting monumental structures because base isolation method does not intervene with

the super structure and has very little effect on the architectural characteristics of the

building. The Oakland City Hall (Figure 3.5.3) and the San Francisco City Hall are some

of the buildings rehabilitated with this method.

63

Figure 3.5.3: Los Angeles City Hall rehabilitated with base isolators (Guerreiro et al.

2006)

3.5.2 Energy Dissipation Devices

The seismic vulnerability of the structure can be effectively reduced by the use of seismic

energy dissipation devices. The use of seismic energy dissipation method increases the

capacity to dissipate energy that has been transmitted to the structure by the ground

moment thereby reducing the seismic effect on the structure. If the system is not

equipped with any energy dissipation devices, energy has to be absorbed by deformation,

elastic or inelastic, of the structural elements. When the structural elements capacity is

not sufficient to accommodate the deformation demand, this can cause fail of the

structural element. By using energy dissipation devices the amount of energy that has to

be absorbed by the structural elements can be reduced by effectively dissipating the

energy. With energy dissipaters, the amount of energy that must be absorbed by the

structure can be controlled and the damage can be limited (Guerreiro et al. 2006).

64

The energy dissipation devices are devices precisely designed to dissipate large amount

of energy without deterioration, and can be distributed in three major classes: viscous

dampers, hysteretic dampers and viscoelastic devices. The hysteretic devices behavior is

based on the plastic deformation capacity of metallic elements, usually steel elements. In

these systems the force depends on the deformation and the control parameters are the

initial stiffness, the yield level and the stiffness after yielding (Guerreiro et al. 2006).

Even if the structure is fitted with energy dissipation devices, these dampers do not

provide any resisting force and the structure has to resist all the lateral loads. A typical

fluid viscous damper has a central piston that strokes though a viscous fluid filled

chamber. The fluid moves through the orifices in the piston head and dissipates energy

and also creates a force that resists the motion of the damper. To ensure proper fluid

performance and stability, silicon-based fluids are used.

A typical arrangement of a viscoelastic damper consists of viscoelastic layers bonded

with steel plates as shown in Figure 3.5.3. A frame structure is needed to install a energy

dissipation device on a structure. So this method is applicable to steel frame structures

and reinforcement concrete moment resisting frames (Guerreiro et al. 2006).

Figure 3.5.4: Arrangement of a viscoelastic damper (Guerreiro et al. 2006)

65

Chapter 4

REHABILITAION FOR BLAST

Most of the buildings are designed without considering blast loading, in wake of the

unexpected events, like recent terrorist attacks, design for blast loading is gaining gradual

attention in the structural community. This chapter summarizes some of the methods that

are currently in use in blast resistant design.

4.1 POLYMER RETROFIT OF UNREINFORCED MASONRY WALLS

Strengthening the unreinforced, non-load bearing concrete partition walls is one of the

primary focuses in the recent years due to the following reasons

1) the frequent use of unreinforced concrete masonry partition walls in building that

generally have high occupancy

2) The susceptibility of these components to fragmentation even in case of a low

blast pressure.

There are several methods to strengthen the unreinforced concrete masonry walls. Some

of the retrofit material includes carbon fiber laminates, aramid composite fibers, etc.

These methods have demonstrated the ability to increase the strength and ductility of

unreinforced concrete masonry to a large extent, but feasibility of wide spread application

of these products is challenged by the development cost, methods of applying these to the

structure. Connell (2000) conducted three full scale tests to determine the effectiveness of

polymers for retrofitting structures for blast loading.

To over come constrains of the above materials experiments were conducted using spray

on polymers. Three full scale experiments were conducted to determine the potential of

these materials in improving the blast resistance of the unreinforced concrete masonry

walls. These tests were conducted to determine the electrometric polymer application

process, measure the deflections at critical wall locations, measure the internal and

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external pressure created by the blast, failure modes and determine the effectiveness and

the level of protection by the electrometric polymer retrofit (Davidson et al. 2004).

In each of these tests several wall panels were tested for blast loading. Some of these

were coated with polymer and while others were not, to evaluate the effectiveness of the

polymers. The wall samples were tested in reusable reaction structures designed to

withstand blast loads. Blast loads were then applied to the panels by detonating explosive

charges from a stand off distance. Instruments were fitted in the structure to measure the

displacements, acceleration and pressure.

4.1.1 Selection of Retrofit Material

In the initial stages, 21 prospective retrofit materials were considered. Seven of the

materials were extruded thermoplastic sheet materials, 13 were spray on polymers and

one was a brush on material. All of the prospective polymers possessed temperature and

ultraviolet stability, flame resistant and were cost effective. To determine the structural

properties of all the polymers MTS load frame tests were conducted at a loading rate of

8.38 mm/s. 4.1. 1 shows the average of the results obtained in the tests for all the groups

(Knox et al. 2000).

Table 4.1.1: Average tensile strength properties obtained from the tests (Davidson et al.

2004)

Application (number of polymers tested)

Secant modulus of elasticity

Elongation at rupture (%)

Maximum tensile strength

Extrusion (7) 113,000 kPa (164,000 psi)

52 55,800 kPa (8,100 psi)

Spray-on (13) 78,500 kPa (11,400 psi)

109 9,650 kPa (1,400 psi)

Brush-on (1) 6,890 kPa (1,000 psi)

25 5,510 kPa (800 psi)

The extrusion polymers were stiffer and stronger than the other groups, but envision

retrofit approach of creating protective shells with in the occupied space made extrusion a

difficult choice to implement. Hence the extrusion thermoplastic polymers are eliminated.

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The brush on polymers were observed to be very weak, brittle and had very long cure

times, hence these are eliminated (Knox et al. 2000).

Spray-on Polyurea was select for the retrofit due to its strength, flammability and cost.

These polymers have many applications ranging from marine application to forming the

lining of food and storage tanks. Material test were conducted on the selected polymer

and the results from the tests are shown in Tables 4.1.2 and 4.1.3.

Table 4.1.2: Properties of Polyurea (Davidson et al. 2004)

Property Measured Value Modulus of Elasticity 234,000 kPa (initial); 165,000 kPa (secant)

[ 34,000 psi (initial); 24,000 psi (secant)] Elongation at rupture 89% Stress at rupture 13,900 kPa( 2,011 psi) Maximum Tensile Strength 14,000 kPa ( 2,039 psi) Toxicity (according to the manufacturer) Non toxic after curing Flame Test (ASTM D635) ATB=infinite, AEB= 19 mm

Table 4.1.3: Material properties obtained for the tests (Davidson et al. 2004)

Sample Maximum Tensile Strength

Elongation at maximum tensile strength

Maximum Elongation (%)

Secant Modulus

Toughness

Polyurea-A 12,700 kPa ( 1,840 psi)

46.4 53.6 180,000 kPa ( 26,100 psi)

5,830 (kPa.X mm/mm) 846 (psi x in. /in.)

Polyurea-B 14,100 kPa ( 2,040 psi)

73.7 83.5 167,000 kPa ( 24,200 psi)

10,100 (kPa x mm/mm) 1,159 (psi x in. /in.)

Polyurea-C 13,200 kPa ( 1,920 psi)

88.6 94.4 152,000 kPa ( 22,000 psi)

10,500 (kPa x mm/mm) 1,522 (psi x in. /in.)

MEAN 13,300 kPa ( 1,930 psi)

69.6 77.2 166,000 kPa ( 24,126 psi)

8,790 ( kPa x mm/mm) 1,276 (psi x in. /in.)

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Figure 4.1.1: Test setup for testing the effectiveness of the spray-on polymer retrofitting

method (Davidson et al. 2004)

4.1.2 Test Procedures

Three tests were conducted to evaluate the effectiveness of the spray on polymer retrofit

technique. In each of the test there was an explosive charge positioned away from a

masonry wall that was built in a highly reinforced concrete structure. Figure 4.1.1

illustrates the test setup. Two masonry walls were constructed inside the reaction

structure. Each of the two walls was 2.24 m x 3.66 m and they are separate by a W12X35

section and 19.1mm foaming on both sides of the wall. These conditions were used to

enforce one way bending. The interior and exterior bottom of each section had been

secured using a 76.2 mm x 102 mm x 6.3 mm (3 in. x 4 in. x ¼ in.) angle and the interior

top had been secured using the same angle and the exterior top had been secured using a

4.88 m x 0.305m x 6.4 m (16 ft x 12 in. x ¼ in.) steel plate.

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One layer of the spray on polymer had been used on most of the walls on the interior and

on the interior and exterior on one wall in test three. The polymer was overlapped in to

the surrounding reaction structure. “Control walls” without the polymer application have

been provided to act as a measuring gauge for the retrofitted walls (Davidson et al. 2004).

4.1.3 Instrumentation

Pressure, acceleration and deflection experienced by the wall were measured by pressure

gauges, single axis accelerometers and laser deflection gauges. The setup of all the

instruments mounted on the walls is shown in Figure 4.1.2.

Figure 4.1.2: Instrumentation plan (Davidson et al. 2004)

All reflection pressure gauges were mounted in a pipe and were suspended from the top

supported from the reaction structure in front of the wall panels. Accelerometers were

attached to the interior side of the polymer retrofitted walls. Laser deflection meters were

placed at the center of both the walls.

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4.1.4 Test Results

Test 1 Results

The typical illustration of the reflected pressure obtained from the test 1 is shown in the

Figure 4.1.3. The maximum values of pressure, deflection and acceleration are listed in

the Table 4.1.4. The peak pressure observed in the experiment is 393 kPa. The laser

deflection gauge indicated that both the walls moved inward by 184mm. the control wall

completely collapsed while the wall with the retrofit remained intact (Figure 4.1.4).

Figure 4.1.3: Test 1 reflected pressure: gauge R1 and R2 (Davidson et al. 2004)

Table 4.1.4: Gauge measurements obtained from Test 1 (Davidson et al. 2004)

Gauge ID Type Measured R1 Reflected Pressure/impulse 393 kPa/1,460 kPa ms

(57 psi/212 psi ms) R2 Reflected Pressure/impulse 362 kPa/1,380 kPa ms

(52.5 psi/200 psi ms) R3 Reflected Pressure/impulse 303 kPa/1,120 kPa ms

(44 psi/163 psi ms) F1 Reflected Pressure/impulse 186 kPa/551 kPa ms

(27 psi/80 psi ms) L1 Laser Deflection 184 mm (7.25 in.) L2 Laser Deflection 184 mm (7.25 in.) A1 Accelerometer 379g A2 Accelerometer 379g A3 Accelerometer 444g

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Figure 4.1.4: View of the damaged walls after test 1 (Davidson et al. 2004)

Test 2 Results

Since the retrofitted wall didn’t collapse in test one. The charge of the explosion was

doubled and the stand off distance was reduced by 14 percent. The instrumental set up is

typically the same as the test1. Due to the increase in the charge and the reduction in the

stand off distance the impact on the wall panel in the test 2 was tripled compared to test1.

The reflection pressures obtained from the test 2 are listed in Table 4.1.5. The values

obtained from laser deflection gauge and the accelerometers were not usable.

Both the control wall and the retrofitted wall panel were destroyed. The control wall

completely disintegrated. The retrofitting polymer held much of the wall together, but the

wall sheared from its supports due to the extreme energy imparted by the blast. As the

wall flexed during the blast the polymer ripped at the height wise center of the wall. The

wall then fell on top of itself in two pieces (Davidson et al. 2004).

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Gauge ID Type Measured

R1 Reflected Pressure/impulse 1,100 kPa / 289 kPa ms

( 159 psi / 419 psi ms)

R2 Reflected Pressure/impulse 1,320 kPa / 1,440 kPa ms

( 192 psi / 209 psi ms)


( 238 psi / 398 psi ms)


( 174 psi / 321 psi ms)

F1 Free Field Pressure/impulse 455 kPa / 937 kPa ms

( 66 psi / 136 psi ms)

F2 Free Field Pressure/impulse 27.6 kPa / 207 kPa ms

( 4 psi / 30 psi ms)

Test 3 Results

Two additional 3.05 m x 3.05 m (10ft x 10ft) wall panels in cubicles were used in this test

as shown in the Figure 4.1.1. A total of 4 wall panels retrofitted with different schemes

are tested. Four unreinforced concrete masonry walls are housed in the reaction structure

(two wall panels) and cubicles (one wall panel each). The left wall panel of the reaction

structure was retrofitted with 9.5 mm (3/8 in.) layer of polymer and the polymer

overlapped the roof and floor slabs by 0.305 m (12 in.). The right panel in the reaction

structure was retrofitted with 3.2 mm (1/8 in.) polymer layer on both the interior and

exterior faces of the wall with a overlap pf 0.305 m (12 in.). Both the walls in the cubical

were retrofitted with 3.2 mm (1/8 in.) polymer layer on the interior with different

overlaps. One of the cubical was provided with an overlap of 0.152 m (6 in.) and the

other was provided with 0.305 m (12 in.). Each panel is allowed to move freely on the

sides to simulate a one-way flexural response. The explosive charge used was the same as

that used in the test 2. The stand of distance was increase by 30% compared to test 1

(Davidson et al. 2004).

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The instruments used in this test were five reflected pressure gauges (R1, R2, R3, R4,

R5), one filed pressure gauge (F1) and four laser deflection meters (L1, L2, L3, L4). All

the laser deflection gauges were mounted as described earlier and as shown in Figure

4.1.2. A free-Field pressure gauge (F1) was located at the stand off distance from the

charge as shown in the Figure 4.1.5. Three of the reflected pressure gauges are mounted

on the reaction structure. R2 was fitted in front of the steel section dividing the wall panel

1.83 m (6 ft) off the ground. R1 was fitted at the center of the left panel 1.83 m (6 ft) off

the ground. R3 was fitted at the center of the right wall 1.83 m (6 ft) off the ground. Two

reflected pressure gauges were mounted on the cubicles: on at the center of the left

cubical 1.52 m (5 ft) off the ground and other center of the right cubical 1.52 m (5 ft) off

the ground.

All the wall panels in test 3 sustained severe damage. The polymer retrofit helped to hold

the walls intact and prevented the debris from fragmenting. Figure 4.1.5 show the damage

to the wall panels in the reaction structure. Figure 4.1.6 show the damage sustained by the

wall panels on the cubicles (Davidson et al. 2004).

Figure 4.1.5: Test 3 Result: Wall panels on reaction structure (Davidson et al. 2004)

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Figure 4.1.6: Test 3 Result: Wall panels on test cubicles (Davidson et al. 2004)

Although the polymer coating was provided on the inside and the outside on one of the

walls in the reaction structure, not sufficient enhancement in protection is provided for

the additional cost. The initiation of tearing on the interior of the wall was shown in

Figure 4.1.8. The reflection pressures obtained from the test 3 are listed in Table 4.1.6.


Gauge ID Type Measured R1 Reflected Pressure/impulse 409 kPa / 1,560 kPa ms

( 59.4 psi / 227 psi ms) R2 Reflected Pressure/impulse 479 kPa / 1,830 kPa ms




( 69.1 psi / 218 psi ms) F1 Free Field Pressure/impulse 87.5 kPa / 655 kPa ms

( 12.7 psi / 95 psi ms) L1 Laser Deflection 239 mm (scratch gauge) (9.4 in.) L2 Laser Deflection 198 mm (scratch gauge) (7.8 in.) L3 Laser Deflection 125 mm (scratch gauge) (4.9 in.) L4 Laser Deflection 140 mm (scratch gauge) (5.9 in.)

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Figure 4.1.7: Tearing of the polymer on the inside from Test 3 (Davidson et al. 2004)

4.1.5 Summary

These tests indicate that spray on polymer technique evaluated here is an effective

method to improve the response of unreinforced concrete masonry walls to blast loading.

Although these tests have shown results that indicate the effectiveness of the tests, the

failure mechanism depends on the peak pressure and duration of the pressure and is not

thoroughly understood. The failure mechanism is also dependent on the support

conditions.

4.2 REHABILITATION OF STRUCTURE AFTER A GAS EXPLOSION

An explosion detonated in a gaseous medium will give rise to a sudden increase in the

pressure in that medium from the ambient pressure to the peak incident pressure of the

explosion. After the explosion a shock wave travels from the blast location at speeds

exceeding the speed of sound. The gas molecules travel at a speed less than the pressure

wave. The pressure and the temperature decreases as the distance is increased. When a

blast occurs inside a building, in a confined space, the hot gases cannot freely mix with

enough cool air. When the pressure wave impacts on a rigid surface, it instantaneously

develops a reflected pressure on the surface which in value is more than the incident

pressure.

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When the peak incident pressure is between 8.3-6.2 kN/m2, the damage to the building is

minimum up to about 5% of replacement cost and the personnel in the building may be

injured by broken glass and building debris. When the peak incident pressure is

15.8kN/m2, the damage to the building is expected to be about 20% of replacement cost

and the personnel in the building may suffer temporary hearing loss and injury from

secondary blast effect. When the peak incident pressure is 24kN/m2, severe damage to the

building can be expected and the injury to the personnel in the building will be of a

serious nature (1% eardrum rupture). When the peak incident pressure is 55.3 kN/m2, the

building damage approaches total destruction, and the personnel in the building will

sustain serious injury. When the peak incident pressure is 82.7kN/m2, building will

sustain severe structural damage approaching total destruction and personnel in the

building can sustain severe injury or death from direct blast. When the peak incident

pressure is 186kN/m2, the building will be completely destroyed and there will be death

of the personal in the building due to the direct action of the blast (Bob 2004).

4.2.1 Rehabilitation Method

Most buildings are not designed to withstand blast loadings. The rehabilitations scheme

for the deterioration structural member depends on the material.

For RC structures, rehabilitation work is carried when the damage is considerably low

and demolished when the rehabilitation costs are high. Repair is done for surface

deteriorations, cracks, damage resulting from improper casting and reinforcement

corrosion. In RC structures increase in stiffness and ductility is achieved by coating the

beams, columns and joints. These coating agents can be RC jacketing, Steel jacketing,

carbon fibers, etc. In some case it is necessary to transform the existing structure

completely. Some of these techniques are:

1) Adding steel bracing to RC structure

2) Infilling of frame holes with reinforced masonry and reinforced concrete.

In steel structures repair involves working on a damaged structure and restoring its

formal structural efficiency. On the other hand, strengthening involves modifications to

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the structural elements or the structure as a whole to improve the structural performance

(Bob 2004).

Rehabilitation of structural elements consists of the following:

1) Adding cover plate connection modifications to the lower and/or the upper sides

and adding bolted brackets to the existing welding connections.

2) Gutting, involves removal of internal elements of the structure and adding new

ones with different properties.

3) Adding new structural elements within the original building.

Masonry structures without reinforcements can be rehabilitated by:

1) Erection of RC cores at appropriate distance.

2) Masonry lining with reinforced concrete

3) Interlocking of masonry walls at corners and crossings.

Conventional methods of rehabilitation may lead to inconvenience and the methods prove

to be costly and the improvements achieved after the retrofit is some times are not as

expected. A typical approach to improve the response of an element to blast loading is by

increasing the structural mass, this leads to seismic problems. Carbon fiber reinforcement

plastics (CFRP) provide a solution to this problem. Structural elements designed only for

gravity load and then retrofitted with CFRP, demonstrated ability to withstand negative

loads and also displayed high ductility. This solves the problem by improving the

structural elements resistance and effectively the mass remains unchanged. Crowford (et

al. 2001) demonstrated a full scale testing of CFRP vertical strips and CFRP horizontal

wraps. CFRP vertical strips and CFRP horizontal wraps were designed, applied and

tested with a blast load full scale field test (Bob 2004).

4.2.2 Case Study of a Structure Damaged by Gas Explosion

In early December 2002 an explosion occurred in a flat in the town of Timisoara,

Romania. The building is a 5 story with 100 flats, and the explosion happened in the

second flat on the fourth floor. The explosion occurred due to a leak in a propane gas

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bottle, the gas accumulated and when the inmates switched on a light in the middle of the

night the gas ignited and caused an explosion. The explosion completely damaged the

concrete walls, concrete floor, windows and doors of the flat. It also caused severe

damage to the other structural members (Bob et al. 2003).

The plan of the building is shown in the Figure 4.2.1. This building is built in 1976 and

the plan dimensions are 43.55 x 14.75 m with a sub-basement and a story height of

2.72m. the building consists of vertical structural member built with longitudinal

reinforcement and transverse reinforcement concrete walls of width 30 cm for concrete

facing the slabs and 15 cm for interior panels. Prefabricated slabs are used in this

structure. The reinforcement of the prefab slabs is 6mm bars at 15 cm c/c over the shorter

span and 6mm bars at 20 cm c/c over the long spam.

Figure 4.2.1: Typical floor plan of the structure (Bob 2004)

4.2.3 Building Damage Assessment

1) The flat directly affected by the blast.

All transverse walls were completely destroyed (Figure 4.2.2)

The concrete facing the slab was moved 55mm from its original position by the

force of the blast (Figure 4.2.3)

Both the bottom and the top slabs showed a maximum deformation of 280-300

mm (Figure 4.2.4).

All the doors and windows were damaged.

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Figure 4.2.2: View of the damaged transverse walls (Bob 2004)

Figure 4.2.3: View of the damaged floor-wall connection (Bob 2004)

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Figure 4.2.4: View of the damaged RC floor (Bob 2004)

2) The surrounding structural members affected by the blast

The concrete facing slabs directly above and below the affected flat were moved

from the original position by approximately 10mm.

The Reinforce concrete floor of the neighboring flat had a maximum deflection of

50mm.

The floors and the transverse walls from1st to 5th floors on one side of the

corridor surrounding the site of explosion were damaged by the shock wave (Bob

2004).

4.2.4 Rehabilitation Scheme

The rehabilitation scheme was selected on bases of technical and economical advantages:

Safe behavior under seismic action.

Slight change of general stiffness of the structure.

Short period of rehabilitation.

Low cost of rehabilitation.

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Figure 4.2.5: strengthening with RC Coating (Bob 2004)

The rehabilitation for this damages structure has been done for the following structural

elements.

1) New concrete floors were added in the levels 3, 4 and 5 with the same geometric

and reinforcement characteristics as of the original members.

2) New reinforced concrete walls at levels 4 and 5 were added with the same width

as of the original members.

3) Local strengthening was carried out by adding a 5cm RC coating on all sides of

columns ( 20x60 cm columns in the corridor and 20x30cm columns in the

facade), longitudinal beams at all levels and vertical elements from ground to the

roof on each side of the corridor (Figure 4.2.5).

4) CFRP (sika wrap) was used to rehabilitate 10 damage transverse walls and 12

damaged floors (Figures 4.2.6, 4.2.7, and 4.2.8) (Bob 2004).

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Figure 4.2.6: strengthening of walls with CFRP (Bob 2004)

Figure 4.2.7: Installation of CFRP to strengthen floors (Bob 2004)

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Figure 4.2.8: Strengthening of floors with CFRP (Bob 2004)

These modifications were chosen in accordance with the condition of the building.

Instead of removing the severely damaged structural member new members were added

to the structure. The coating on the columns was necessary since the columns were

severely damage and this ensures connectivity and stability of the vertical members. The

use of CFRP to strengthen the slightly damages members of the structure made it easy to

install and a short time of erection. The rehabilitation scheme followed is shown in the

plan view in Figure 4.2.9 and 4.2.10. The strengthening of the floor by CFRP was shown

in the Figure 4.2.11 (Bob 2004).

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Figure 4.2.9: Strengthening with RC sheets, columns and longitudinal beams (Bob 2004)

Figure 4.2.10: Local strengthening of columns and beams with RC jacketing (Bob 2004)

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Figure 4.2.11: Strengthening of floors with CFRP (sika wrap) (Bob 2004)

4.2.5 Summary

The Damage to the structural members in the building is dependant on the position, the

members that are in direct contact with blast were completely damaged and the ones near

by are slightly damaged. This strengthening solution has been considered to obtain

technical and economical advantages.

4.3 BLAST RESISTANT DESIGN OF COMMERCIAL BUILDINGS

There are numerous commercial office and retail buildings throughout the United States

that are vulnerable to terrorist attack. Here we are examining a prototypical commercial

office building and examining the deficiencies and vulnerabilities of the system for blast

loading and recommending modifications that improve the performance of the building

against blast loading.

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The building chosen for this study is a cast-in-place reinforced concrete eight story

structure. Though this is a very commonly found structure, it has very high vulnerability

for blast loadings. Therefore, this structure was chosen to identify the vulnerabilities and

improve it for blast loads (Ettouney et al. 1996).

The eight story structure considered in this study is show in the Figure 4.1.1. The

structural system of all the eight stories and the slab is typical of reinforced concrete flat-

slab construction. The columns, slabs and beams all are cast-in-place concrete. In this

structure columns are spaced 9 m center to center and the typical floor height is 3.9 m and

the first floor height is 6 m. Spandrel beams are provide on the out side edge of the

building to support the facade. The lateral loads are resisted by the shear wall around the

elevator chambers. This building has been designed to resist wind loads and seismic loads

(Ettouney et al. 1996). The seismic load specification confirm to the uniform building

code (UBC 1991). This building occupies a complete city block and has public access on

all four sides with a surface parking lot at the rear of the building. There is also a loading

dock at the rear of the building. One of the distinctive features of this office building is a

two story atrium in the front of the building, faced with exterior glass on the perimeter

and with open in the interior. The perimeter of the atrium is ringed by a spandrel beam

that supports the outer edge of the slab. Above the atrium, the column spacing of 9m

center to center is restored by the transfer girder (Ettouney et al. 1996).

Figure 4.3.1: Elevation view of the building (Ettouney et al. 1996)

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Figure 4.3.2: Plan view of the building (Ettouney et al. 1996)

Figure 4.3.3: Isometric view of the building (Ettouney et al. 1996)

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4.3.1 External Treatment

The most important factors that influence the blast environment are

1) Stand off distance (The distance of the charge from the object).

2) Bomb’s charge weight

Since the charge of the bomb can not be controlled, the only factor anyone can control is

the standoff distance. Irrespective of the charge of the bomb maximum possible stand off

distance must be provided around the entire perimeter of the building in every case so as

to reduce the effect of the explosion. The main factor stand off distance depends is the

site selection.

For the building under consideration the only area that can be controlled is the side walk

around the building to limit the stand off distance. Since there is not much stand off

distance this building is highly vulnerable to hand thrown bombs and car bombs. The

most directly effected structural members are the columns beams and the atrium in the

first floor and the facade. Since these members are highly vulnerable special attention is

required (Ettouney et al. 1996).

4.3.1.1 Stand off Distance

Maximum stand off distance, within which explosive laden vehicles may not penetrate,

must be provided and guaranteed. As we all know greater the stand off distance greater is

the dissipation and the building experiences lower pressure and impulse. Following are

the recommendations that can increase and improve the effectiveness of the stand off:

1) Use anti-ram bollards or large planters, placed all around the building. The

bollards must be design to resist very impact loads. While designing the bollards

care must be taken on the maximum speed that a vehicle can attain, site

conditions govern the maximum speed that can be attained by a vehicle. And

while designing the connections for the bollards, these must be designed for the

maximum impact load that can be applied on the bollards. When there are design

restrictions on the bollard connection design care must be taken to reduce the

maximum speed attained by applying field modification.

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2) The public parking lot at the back of the building is a vulnerable point to plant a

car bomb. So, this parking lot must be secured and all the vehicles that enter the

parking lot must be cleared, i.e. employee vehicles or vehicles visually examined.

If possible parking must be eliminated on the near side of the parking lot.

3) Street parking typically on the near side of the building has to be eliminated. This

reduces the chances of parked car bomb. Normally, the city gets high revenue

from street parking, so the owner of the building has to pay for the loss of

revenue.

4) Additional stand off distance can be attained by removing one lane of traffic

around the building and using this space as a walkway. This has to be approved

by the city officials and could also cause traffic problems.

5) An additional measure to improve the effectiveness of the stand off distance is to

remove the street parking on the other side of the road. Even though this will not

improve the stand off distance, this will decrease the chances of parked car

bombs. This will not reduce the chances of “park and run”, drive-by and suicide

bombs. Unfortunately, in the case of the Oklahoma City bombing, a truck laden

with explosives is parked for approximately for 2 minutes and still it dint grab the

security officer’s attention.

It must be understood that increasing the stand off distance is very effective in case of a

small charge. When the bomb has a very high charge, an increase in stand off by 9 or feet

will not be enough, the blast forces may overwhelm the structure even after the increase

(Ettouney et al. 1996).

4.3.1.2 Lower Floor Exterior

The architectural design of the building used glass facade on all sides of the building.

Unless this has been replaced with reinforced concrete walls, the lower level structural

member and their connections will suffer heavy damage. This will also cause severe

damage to inhabitants of lower floors. In general three cases of charges and the corrective

action are discussed below:

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1) To protect against a small charge, a 300mm thick wall with 0.3% steel double

reinforcement in both direction is required.

2) To protect against a intermediate size charge, a 500mm thick wall with 0.5%

reinforcement in both direction is required.

3) In case a large charge, the blast will breach any reasonable sized wall and damage

the lower floor structural members. So protect against a large charge, precautions

must be taken and adjustments must be made for the whole structure (Ettouney et

al. 1996).

4.3.2 Glazing

Glazing is the first weak link when blast loading is considered. For any reasonably sized

bomb, all the glazing will shatter especially on the side facing the bomb. Commonly used

annealed glass behaves very poorly when loaded drastically. The failure mode for the

annealed glass will create large sharp objects resembling knifes and shatter them all

around inside the structure and cause heavy injuries and causalities. It has been observed

in the previous blasts that glass shattered will cause heavy damage and that is one of the

important factors.

While typically annealed glass can only withstand 14KPa (9psi) of blast load there are

other alternative glazing materials which can resist a greater amount of blast load and can

perform better in some case of the blast loadings. Thermally Tampered Glass (TTS) and

Polycarbonate lay-ups can be made in to sheets of one inch thick and can withstand

higher blast loads than the annealed glass. Unlike annealed glass, TTS normally breaks in

to rock-slat sized pieces and infringe less injury to the inmates. TTS can normally

withstand 200 – 275 KPa (30 – 40psi). TTS is used in the side and rear windows of the

automotives. Failed Polycarbonate lay-ups remains in one large piece and will also cause

heavy damage to the inmates same like annealed glass. Polycarbonate lay-ups is normally

used in the windshields of cars. Care must be taken when design of the glass is carried.

While designing the connectors for the design of glass, the connectors must be designed

in such fashion that full strength of the glass has been achieved before any failure of the

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connectors. If the connectors fail before the glass, the glass as a whole will fly and

infringe heavy damage (Ettouney et al. 1996).

4.3.3 Facade and Atrium

As the exterior of the building is the first defense against blast loading, how the facade

responds to blast loading is an important factor. The facade consists of the glazing and

the exterior walls. Realistically glazing is pressure sensitive and is the first building

component that will fail in presence of blast loading. Although there are methods to

strengthen the exterior walls, the number of methods available to strengthen glass and

windows are limited. Windows will break during the blast and the pressure wave will

enter the building. There is direct correlation between the degree of fenestration and the

amount of blast that is allowed to enter the occupied space. Limiting the amount of

fenestration will also limit the blast effects. Atriums are common on prestigious office

buildings. These atriums will provide the buildings with a grand look and also allow the

natural light to enter the lobby area. And also provide excellent functioning spaces and

balcony elevator lobby. Atriums are inviting target for the reason that, the damaged glaze

will expose all the structural elements and this may cause a lot of damage to the structural

system. The building under consideration has a big window at the entrance, this window

cannot be strengthened to withstand the blast loadings, so the shattered fragments from

this window and the glaze will cause a potential hazard to the inmates and this will also

allow the pressure wave to enter the structure (Ettouney et al. 1996).

4.3.3.1 Exterior of the Atrium

The exterior of the atrium is very important. The window provided at the entrance and the

glazing must be strengthened to resist small charges so as to protect inmates or the

exterior glazing can be replaced with reinforced concrete walls and protect the structural

system from moderate-sized charge (Ettouney et al. 1996).

4.3.3.2 Interior of Atrium

Almost all blast will generate a pressure greater that 14 KPa (2psi) on the external facade

of the structure destroying the glazing and letting the blast wave enter the building. The

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blast wave will dissipate energy both with distance and reflection off of the internal

surfaces. Depending on the charge there will be a height where the blast wave intensity

has been reduced to 70 – 100 KPa (10-15psi), a magnitude where human fatality is not

likely. At this elevation the blast wave that enters the structure thought the shattered glass

will reduce sufficiently to pose a reduced threat to the occupants. In case of a moderate

sized charge explosion outside the glass facade will cause a extensively high blast wave

intensity of 200 KPa (30psi) over the entire height of the building. The shock wave

entering the structure will damage the partition walls and pose danger to the occupants.

The only way to protect the structural elements is by strengthening the balcony parapet,

spandrel beams and exposed slabs to resist blast loads.

4.3.4 Floor Slabs

When the structure is subjected to blast loads, the flat slab construction is subjected to

large dynamic pressure load. The softening of the moment resisting capacity of the slabs

will reduce the lateral-load-resisting capacity of the system. When the moment carrying

capacity of the slabs at the columns is lost the ability to transfer the loads to the shear

walls is also reduced and this will weaken the structure severely (Ettouney et al. 1996).

These are some of the possible modes of failure:

Figure 4.3.4: Failure mechanism of the flat slab (Ettouney et al. 1996)

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Figure 4.3.5: Effects of blast loading on columns (Ettouney et al. 1996)

Figure 4.3.6: Lateral load carrying mechanism and effects after the blast (Ettouney et al.

1996)

1) The slab by itself can undergo localized failure as shown in Figure 4.3.4.

2) The loss of the connection between the columns and the slabs will increase the

unsupported length of the column and may lead to the buckling of the column.

This has been depicted in the Figure 4.3.5.

3) The lateral load carrying system consists of the columns, shear walls and the slab

that transfers the lateral loads. This system can be damaged to such an extent that

the structure as a whole is unstable for lateral loads. This has been depicted in the

Figure 4.3.6.

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Figure 4.3.7: High and Low vulnerability locations (Ettouney et al. 1996)

Figure 4.3.8: Flat slab improvements (Ettouney et al. 1996)

To avoid such calamities, following improvements must be considered.

1) More attention must be taken while designing the exterior bays and lower floors

which are very vulnerable to blast loads. The graphical representation of this is

shown in the Figure 4.3.7.

2) Spandrel beam which are not mandatory in the structure must be provided and

tied together to improve the response of he slab edges.

3) In exterior bay and lower floors, column heads and drop panels must be used to

enhance the punching shear resistance as shown in the Figure 4.3.8.

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4) If vertical clearance is a problem, shear heads embedded in the slab will improve

the shear capacity and also improve the transfer of the moments from the slab to

the column.

5) The ductility demands and the shear capacity to resist multiple load reversals

make it a ideal place to provide a beam to span over the critical sections of the

slab.

6) Bottom reinforcement must be provided continuously through the column, this

prevents brittle failure (Ettouney et al. 1996).

The slab must be designed for punching shear as a collapse of the slab may cause a

progressive collapse. Hawkins and Mitchell (1979) showed that a punching shear failure

in an internal column is more likely to cause a progressive collapse than a damaged

external column due to a car bomb.

4.3.5 Columns

For typical building columns are designed for gravity loads and no special consideration

is taken to design for blast loading. In case of blast loading the column has to be design to

withstand large lateral deformations and have high ductility. Incase of blast loading

depending upon the distance of the charge, the characteristics of the loading of the

column can be summarized in to two different scenarios:

1) A building located at a distance more than the perimeter proximity of 30m from

the charge will experience a low pressure uniform distributed all over the facade.

2) Building that are situated less than 30m from the curb are more prone to be be

exposed to more localized, high intensity blast loads.

Most of the columns that are designed for gravity loads must be taken in to consideration

for improvements to resist blast loads. As shown in the Figure 4.3.9, when an external

column is subjected to blast loading, it experiences high blast pressure. This blast

pressure will result in heavy bending of the column in addition to the axial load of the

column. To withstand the axial load and the lateral deformation due to the blast load, the

column has to be designed with sufficient ductility (Ettouney et al. 1996).

96

Figure 4.3.9: Direct Lateral loading of column (Ettouney et al. 1996)

Figure 4.3.10: Uplifting on the columns (Ettouney et al. 1996)

97

The blast pressure that enters the building through the shattered windows and openings,

will load the underside of the slab and then the upper side, with some time delay. This

time delay in loading the slab may some times cause a net upward force on the slab and

the difference in the pressure on both sides of the slab, May also creates an upward force.

So the slabs must be designed to resist loads opposing gravity. When there is an upward

force on the slabs, even the column may sometimes experience a net tensile force (fig10).

The conventional columns on the building under consideration are not designed for

combined effect of bending and tension and therefore may be prone to damage (Ettouney

et al. 1996).

These are some of the recommendations that improve response of the columns for blast

loading.

1) The potential direct lateral load and the impact from the blast debris on the lower

level column make it important for these columns to have high ductility and

strength.

2) The lower level perimeter columns must be designed to resist high lateral loading

that can be caused due to a blast, as show in the Figure 4.3.9.

3) The column ductility and the strength can be improved by encasing the column in

a steel jacket. This also improves the concrete confinement.

4) The possibility of uplift in the columns has to be considered and if deemed likely,

measures must be taken to design the column for a transient tensile force.

5) For smaller charges, spiral reinforcement in the columns will improve the

confinement of the columns and improve the ductility and strength of the column

to resist the blast.

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Figure 4.3.11: Progressive collapse mechanism (Ettouney et al. 1996)

4.3.6 Transfer Girders

Transfer girder transfers the load for the columns above the atrium to the adjacent

columns out side the atrium. Transfer girder spans the width of the atrium; this allows a

column free architecture for the atrium. All building with transfer girder must be treated

with care when blast loading is considered. A blast load that adversely affects the transfer

girder may trigger a progressive collapse and destroy the whole structure (Figure 4.3.11).

The column connections must support the transfer girder even after in elastic deformation

(Ettouney et al. 1996).

The following recommendation will improve the blast resistance of the transfer girder.

1) Adequate detailing must be used while designing the column connections for blast

loading.

2) If the blast loading exceeds the girder capacity, a progressive collapse analysis

must be preformed.

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Chapter 5

CONCLUSIONS AND RECOMMENDATIONS

5.1 CONCLUSIONS

In summary the following conclusions are drawn from the study.

Wrapping a Beam-Column joint with GFRP will enhance the response of the joint

to seismic loading and transform the brittle behavior to a more ductile behavior.

High strength concrete columns when confined by a stay-in-place FRP

confinement can develop ductile behavior under simulated seismic loads.

The increased confinement requirement of the high strength concrete columns can

be met with the use of FRP stay-in-place confinement. Unlike the discrete nature

of the conventional reinforcement, FRP confinement provides continuous

confinement, covering the whole column face, thus providing better confinement

efficiency.

RC jacketing of columns has many advantages. This method improves the

stiffness of the column uniformly all through the column length. Sand blasting is

an effective technique for interface surface preparation.

Adding shear walls to a structure will effectively increase the lateral stiffness of

the structure and act as the primary lateral load carrying system thereby protecting

the old structural frame from seismic loading. Adding shear walls to a structure

will reduce the undesirably large lateral deflections.

The use of base isolation system will effectively retain the architectural beauty of

old buildings and enhance the response of these building to earthquake loadings.

This is the most suitable method for historical buildings were the architectural

aspects of the building can not be modified.

Providing maximum stand off distance is the best possible solution for protecting

a structure from the effect of blast.

The spray-on polymer approach to strengthen unreinforced masonry wall for blast

loading is quite effective. The peak pressure that the wall can withstand increases

100

many times after the application of the polymer. The polymer layer also helps in

preventing fragmentation of the wall.

The use of Thermally Tempered Glass (TTS) and Polycarbonate lay-ups in place

of annealed glass is a preferred option to prevent shattering of glass after blast and

thus protect the occupants from serious injuries. A summary of various rehabilitation techniques for resisting earthquake and blast loading

are provided in Tables 5.1.1 and 5.1.2. Table 5.1.1: Rehabilitation Techniques for Resisting Earthquake Loading

Technique Comment

FRP wrapping of beam-

column joints

Wrapping beam column joints with FRP laminates

increases the ductile behavior of the joint and the

resistance to earthquake loading.

FRP stay-in-place

confinement for HSC columns

High strength concrete is generally brittle. So the

confinement requirements are very high. Using

conventional confinements with high strength concrete

will create cage congestion. This can be avoided by

providing FRP stay-in-place confinement.

RC-jacketing of concrete

columns

By RC jacketing of concrete columns, significant

increase in strength and ductility of the column can be

achieved. This method improves the stiffness of the

column uniformly all through the length of the column.

Shear walls Shear walls are added to the structure in order to

increase the lateral stiffness of the structure. Soft story

failure is observed in structures with low lateral stiffness

in the lower floors. This can be avoided by adding shear

walls in the lower floors.

Bracing Similar to shear walls bracings also increase the lateral

stiffness of the structure. By adding bracings the lateral

deformation in the event of an earthquake can be

reduced effectively.

101

FRP laminates for walls Ductility of the wall panels can be improved by adding

FRP laminates. This will increase the lateral stiffness

and minimize the deformations.

Base isolators

Base isolators decouple the horizontal movement of the

base from the horizontal movement of the

superstructure. This is an expensive method, but is

useful to maintain the architectural integrity of historical

structures.

Energy dissipation devices Energy dissipation devices are installed in the structure

to effectively dissipate the energy transmitted to the

structure and reduce its deformation.

Table 5.1.2: Rehabilitation Techniques for Resisting Blast Loading

Technique Comment

Increasing stand-off distance Most effective technique but large stand-off distance

may not always be possible for effective functioning of

the building (e.g., building in downtown of a crowded

city, hospitals, etc.)

Spray-on polymer This technique is useful in improving the ductility of the

structural elements. This method is suitable for

unreinforced masonry walls. This helps in preventing

complete shattering of the wall. The peak incident

pressure that the wall can resist can be improved by this

method.

Use of Thermally Tampered

Glass (TTS) or Polycarbonate

lay-ups for glazing

Annealed glass that is typically used for glazing has

brittle nature and can not withstand blasting loading.

Thermally Tampered Glass (TTS) or Polycarbonate lay-

ups can be made into sheets of one inch thick and can

withstand higher blast loads.

Steel jacketing of columns Lower floor columns experience heavy lateral forces

during a blast. So the ductility of the columns has to be

102

increased to improve the column resistance to blast

loading. Jacketing of columns increases the ductility and

strength of the columns.

FRP laminates for walls In this method FRP laminates are attached to the walls,

which increase their ductility. The walls can then

withstand higher peak pressures without fragmentation.

5.2 RECOMMENDATIONS

All the methods discussed in this report have advantages and disadvantages, and some

have applicability limitations. The adaptation of a particular method is also dependent on

economic consideration. Hence a detailed economic comparison between applicable

methods is necessary before rehabilitating a given structure.

The feasibility of using nondestructive evaluation methods to monitor and asses the

condition of the FRP composite bonded to the structure should be considered.

Finite element models should be used to simulate the blast conditions and asses the

structural behavior, thereby facilitating the understating of the fracture mechanism.

A combination of two or more of the aforementioned rehabilitation techniques can be

implemented for more effective energy dissipation and hence obtaining a more ductile

behavior.

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